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Home My WebLink About; Tidal Inlet Risks Agua Hedionda Lagoon; Tidal Inlet Risks Agua Hedionda Lagoon; 1996-07-26Ill m ANALYSIS OF TIDAL INLET CLOSURE RISKS AT AGUA HEDIONDA LAGOON, CALIFORNIA, AND POTENTIAL REMEDIAL MEASURES P li m m p to m, m by Scott A. Jenkins, Ph.D. and Joseph Wasyl FULLY IMPLEMENTED 4 STAGE DREDGE PLAN AGUA HEDIONDA LAGOON. CA CONTOURS IN TOT MCVD +4 to -8: ->. -10. -13. -16 CHANNEL SLOPE - ».4:1CAPACITY: tu,ao c? ttST BASDI RUB: 130.UO CJ SAND CAP IM.T20 cr SLOPE ' 10 1 9EDIHENT KAP OR seerr A mama COKTOLTWC TOTAL DREDGE voumt • Tt&oeo cr SCOTT A Jxracms PU> * JOSEPH WASYL Submitted to: Submitted by: San Diego Gas and Electric Company July 26, 1996 Scott A. Jenkins Consulting 14765 Kalapana Street Poway, CA 92064 m TABLE OF CONTENTS I) STATEMENT OF THE PROBLEM 1 A) Effects of Increased Generating Capacity 1 B) Effects of Sedimentation since Lagoon Construction 6 H) TECHNICAL APPROACH 15 m) TIDAL HYDRAULICS AND PLANT IN-FLOW MODELING 16 A) Tidal Hydraulics Model 16 B) Ocean Tidal Forcing Function 19 C) Plant Flow Rate Forcing Function 28 D) Results for Existing Lagoon Bathymetry 28 IV) WAVE REFRACTION/DIFFRACTION MODELING AND LONGSHORE TRANSPORT RATES 44 A) Wave Forcing File 47 B) Refraction/Diffraction Code and Gridding Systems 49 C) Refraction/Diffraction Results 51 D) Longshore Transport Rates and Updrift Dredge Disposal 60 V) ANALYSIS OF THE RISK OF INLET CLOSURES 62 A) Recharge Rate 62 B) Closure Probability Computations 68 C) Closure Probabilities for Existing Lagoon Bathymetry 70 VI) ANALYSIS OF POTENTIAL REMEDIAL MEASURES TO REDUCE INLET CLOSURE RISKS 79 A) Manipulation of Plant Flow Rates 79 B) Modification of Choke Points 86 C) Reconstruction Dredging 120 VII) CONCLUSIONS 140 REFERENCES 143 Ill m m APPENDIX I: APPENDIX H: APPENDK HI: APPENDDCIV: APPENDK V: APPENDK VI: APPENDIX VH: APPENDIX APPENDDCIX: APPENDK X: m it J* m SOURCE CODE TO OCEANRDS_ NESTED_ GRID MONTHLY PLOTS OF CLOSURE PROBABILITIES FOR EXISTING BATHYMETRY OPTIMAL VELOCITY DISTRIBUTIONS FOR PRESSURE RECOVERY DURING FLOOD FLOW OPTIMAL VELOCITY DISTRIBUTIONS FOR PRESSURE RECOVERY DURING EBB FLOWS STRATFORD STRATFORD CALIFORNIA COORDINATES FOR THE FLOW FENCES OF THE FLAT BOTTOM SOLUTIONS (STRATFORD FENCE) CALIFORNIA COORDINATES FOR THE AUXILIARY FLOW FENCES OF THE VARIABLE DEPTH SOLUTION (STRATFORD BOTTOM) WATERLINES AND EAST BASIN WATER ELEVATIONS FOR STAGE 1 DREDGE PLAN DURING EXTREME SPRING AND NEAP TIDES AND MEAN TIDE CONDITIONS WATERLINES AND EAST BASEST WATER ELEVATIONS FOR STAGE 2 DREDGE PLAN DURBSTG EXTREME SPRING AND NEAP TIDES AND MEAN TIDE CONDITIONS WATERLINES AND EAST BASEST WATER ELEVATIONS FOR STAGE 3 DREDGE PLAN DURESIG EXTREME SPRE«JG AND NEAP TIDES AND MEAN TIDE CONDITIONS WATERUNES AND EAST BASEST WATER ELEVATIONS FOR STAGE 4 DREDGE PLAN DURESTG EXTREME SPRE^O AND NEAP TIDES AND MEAN TIDE CONDITIONS f 1 ANALYSIS OF TIDAL INLET CLOSURE RISKS AT AGUA HEDIONDA LAGOON, CALIFORNIA, AND POTENTIAL REMEDIAL MEASURES By Scott A. Jenkins, Ph.D. and Joseph Wasyl I. STATEMENT OF THE PROBLEM M m Although the lagoon has never suffered an inlet closure since its construction in 1954, a susceptibility to closure conditions has been steadily increasing, particularly over the last te fifteen years. This enhanced susceptibility is due to two intervening factors: ^ 1) Increased plant generating capacity, resulting in greater diversions of lagoon water ^ through plant condenser systems, and ^» 2) 692,000 cubic yards of sediment deposition in the central and east basins, where m maintenance dredging has not been performed. m Both these factors diminish the net volume of water (tidal prism) exchanged with the l» ocean through the ocean inlet during each tide cycle. Consequently, tidal velocities in the ocean pi inlet will more frequently drop below the critical threshold velocity of the native sediments, fc whence the ocean inlet ceases to remain self-scouring. In this way, the inlet has become more M susceptible to closure. m ft A. Effects of Increased Generating Capacity * The plant infall, located as shown in Figure 1, makes Agua Hedionda distinctly P different from natural tidal lagoons in that it functions as a kind of "negative river." Whereas most natural lagoons have a river adding water to the lagoon, causing a net outflow at the ocean P inlet, Agua Hedionda's infall removes water from the lagoon, resulting in a net inflow of water fl through the ocean inlet. That net inflow has two adverse consequences on inlet stability: 1) it W draws suspended sand from the surf zone into the lagoon to form bars and shoals which restrict the tidal circulation, and 2) the net inflow of water diminishes or cancels the ebb flow velocities ? out of the inlet, thereby inhibiting or preventing scour and facilitating deposition of sediments in the inlet channel. Therefore, the plant demand for lagoon water (which scales with user m m f * Ei if fi t » ii ii t i ft t i tt *» ii t i t i f i fi i i ti AGUA HEDIONDA LAGOON, CA GRID CELLS 25 x 25 ft EXTREME POTENTIAL HHW = +4.26 ft NGVD EXTREME POTENTIAL HHW = +4.26 ft NGVD PLANT INFALL DR SCOTT A JENKINS CONSULTING CARTESIAN COMPUTATIONAL GRID SCOTT A. JENKINS PhD & JOSEPH WASYL FIGURE 1: Cartesian computational grid for the tidal hydraulics model. m 3 m demand for power) is a major factor jeopardizing inlet stability. •" The final construction dredging report by Ellis (1954) indicated that the lagoon system *• initially had a mean tidal prism of 55 million cubic feet and a maximum diurnal or spring tidal •» prism of 80 million cubic feet. Since that time, five power-generating units have been brought M on line, increasing the maximum flow rate diverted from the lagoon through plant condenser ** systems to 808,000,000 gallons/day (808 mgd). Plant diversion of lagoon waters at this flow ** rate reduces the net portion of tidal prism flowing out the ocean inlet during ebb flow by ** 27,939,664 cubic feet, or a 50.8% reduction in the original mean tidal prism and a 34.9% «*» reduction in the original maximum spring prism. During high user demand periods, plant flow m rates are typically 635 to 670 mgd (see Figure 2), which reduces the volume of water available ** to flush ocean inlet by 39% to 42% for mean tidal ranges. The designer of the original construction profile for the lagoon was Omar Lillivang. It ** is apparent from the SDG&E file copies of his design report and other correspondence that hemdid not anticipate as many as five generating units and especially not the very large class of *" generating unit such as Encina Unit 4 (306 megawatts) or Encina Unit 5 (346 megawatts).fc Consequently, Lillivang could not have designed for consumption of lagoon water anywhere near ** the present levels that are realized during peak user demand for power, as shown in Figure 2. Lillivang did, however, allow for a safety factor when sizing the lagoon tidal prism. The primary design criteria in the early 1950's was the empirically derived Johnson-O'Brien curve, later recalculated for the U.S. Army Corps of Engineers by Jarrett, as shown in Figure 3. The Q| m design criteria specified by this curve establishes definite limits on the size of the diurnal tidal prism that will maintain an open ocean inlet of a given cross-sectional area. Because the inlet fj dimensions at Agua Hedionda are fixed by rip-rap jetties and channel fortification, there is a lower limit below which the available tidal prism during ebb flow will no longer maintain the ^H w inlet by natural current scour effects. Considering the reduction in net tidal prism due only to the present day operation of five generating units, the design criteria specified by Figure 3 wouldw ^ conclude that the 80 million cubic feet of diurnal prism for the original dredge project is barely ^ adequate. The ocean inlet at Agua Hedionda has about 1,450 square feet of cross-section below « mean sea level. Figure 3 indicates that 58 million cubic feet of diurnal prism is required, which DM W c* t i •» c» * t tfi t> rt n ri ri ri •oa E 1 SDG&E ENCINA POWER PLANT OUTFALL FLOW RATE JULY 27, 1993 - JULY 27, 1994 1000 r- 800 600 400 200 I I • ACTUAL PLANT FLOW RATE IT 0 50 100 150 200 250 300 350 TIME ( DAYS ) FIGURE 2: Plant inflow rate time history for the calendar year July 27,1993 to July 27,1994 INLET AREA DEPENDENCE ON POTENTIAL TIDAL PRISM «• fc* 00 m 1011 M m n _>^^ <~ <fk < IQ:° • I Q «• (->-• L o ™ O •» 5 ^ CO TO9 Z> «• CO fc, Qi•» CU *» Q |^ £— < 1C1 it o E- «f «• « •tn? AREA m2 10 ic2 :o3 l i i ••? t 10J I i — i — ; ;,,;;; i — i — , ; , , ; ; ; i — i — . i . i , ', 1 i • ;/r I t i i I I r i 1 i i i i i i i 1 1 1 1 « 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 ' ' ' ' ' LEGEND ' ' ' ' ' A ATLANTIC COAST - NO JETTIES • ATLANTIC COAST -ONE JETTY 0 ATLANTIC COAST -TWO JETTIES 1 1 1 1 1 I PACIFIC CCASr-ONfc JfcTTY / "" 0 PACIFIC COAST -TWO JETTI65 ]| j/ 1 III ! II «.f * MI t i i ,.r. i ill i X 1 1 l^ .A ' y111 ,n. vfi JK >f ^ 1 */\ 1 1 1 •/, 'Jj 1 /I I 1 -'«X * 1 X ( 1 J ijy? * IX ^k/ttl • V' ,*n i ilfiiil i i >^l"''i|-T3. ^" 1 (Ml 1 1 • IL l^V"7x" f 1 / 1 5H/. if • Mill 1 i t I t /T rf ™ i t i 1 1 l 1 1 1 -II — H — TTTl 1 i IXi'l *lx!S» *"/ Nil I/XI 121III/ • /' V^ &/ •> '/ \ 'I/ ,*/t 1 Xil ...... • . i^-i— ^/-t-^^/; ; ;;;; : u— ' 1 | ~f/: 1 1 1 I VX 1 , '• 'S 1 1 1 1 1 1 1 1 1 .(.S >/ MX 1 .' 1 I ' ' '•$' '<M ^x' i ^' ,^111^ || III ' \ 7 \A if 1 II •• 1 t 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 1 \ ' 1 1 1 t II • , p» 10 i— • '•• ^ ^ <- MINIMUM CROSS SECTIONAL AREA OF INLET BELOW MSL •i 1 1 "• II 1 ' 1 1 —rii 10s ft2 101U TO9 10s o p S CO t Q c_ 10^ 1C6 FIGURE 3: Equilibrium cross section of a tidal inlet as a function of potential prism according to the Johnson-O'Brien curve, from Johnson (1973) and O'Brien (1931). *. 6 «•> during ebb tide is somewhat less than the net of the plant flow rate and the original dredge •* project diurnal prism, «• ""* 80 million cubic ft - 27.9 million cubic ft = 52.1 million cubic ft. «• llf Therefore, the safety factor built into the tidal prism sizing of the original dredge construction m- profile has been completely used up by unforeseen levels of expansion in the plant generating M capacity. This has left no residual safety factor to deal with the additional losses of tidal prism ** due to sedimentation. m ** B. Effects of Sedimentation Since Lagoon Construction *t^^ A decisive question which this report addresses is whether or not the 692,000 ** cubic yards of sediment infilling in the east basins has reduced the available prism sufficientlymto threaten periodic closures of the ocean inlet. Not all of this sediment infilling has produced a direct reduction in tidal prism, because some of that infilling has occurred below the diurnal lower low water level (LLW). Since tidal prism is the volume of water exchanged between the lagoon and the ocean due to tidal action, only the volume contained between the LLW level and the higher high water level HHW contributes to the diurnal prism. Figure 4 gives depth HI m contours based upon sounding data obtained in June 1993 with updates in November 1995. The prominent shoals which have resulted in static displacement of potential tidal prism volume arem w a sand bar in the middle basin of the inner lagoon system, as shown in Figure 5, a pair of islands in the eastern basin, photographed in Figure 6, and a foreset muddy delta formation in P^ff ^ the far eastern end of the east basin (see Figure 7). The volume of water which has been displaced by these features within the diurnal tidal range has resulted in a static loss of tidal «u prism. However, additional losses in prism have also resulted indirectly from shoaling below LLW by reducing the local water depth and thereby increasing the friction acting against the 4w tidal flows. Increased bottom friction retards both flood and ebb flow and prevents water levels ^ within the lagoon from equalling water levels at the ocean inlet. This reduces the tidal range ** inside the lagoon (referred to as tidal damping) which further reduces the achievable tidal prism. IV I* ft it «t ft tt ri ti ti it ti it AGUA HEDIONDA EXISTING BATHYMETRY PER JUNE 1993 SURVEY, UDPATED NOVEMBER 1995 RECHARGE ZONE DEPTH CONTOURS + 4, + 2, 0. -2, -4 ft NGVD -10, -15, -20, -25, -30 ft NGVD 0 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET DR. SCOTT A. JENKINS CONSULTING EXISTING CONTOURS - AGUA HEDIOBNDA SCOTT A. JENKINS PhD & JOSEPH WASYL FIGURE 4: Bathymetry of the existing lagoon system per soundings in June 1993 and November 1995. — -r £ •fc» FIGURE 5: Sand bar in the middle basin as viewed looking east towards the 1-5 bridge."V FIGURE 6: Sand bars in the western end of the east basin as viewed looking west towards the 1-5 bridge. FIGURE 7: Mudflats of the foreset delta formation in the eastern end of the east basin 11 The effects of sediment infilling on reducing tidal prism and increasing closure risk have built up progressively over time, but were accelerated when a recent climate shift ensued with a series of wet El Nino winters beginning in 1978. Increased rainfall and runoff during the winters of 1978, 1979, 1980, 1983, 1986, 1988, 1993 and 1995 (see Figure 8), have caused significant top soil erosion in the surrounding watershed, resulting in large episodic discharges of fine-grained sediments into the east basin of the lagoon. These discharges have created substantial new mudflats throughout the east basin, but particularly in the far eastern end where the Agua Hedionda Creek has deposited the large muddy delta formation, shown in the photograph of Figure 7. There has also been infilling and shoals due to coarser-grained sediments, including fine sand and native beach sand. These shoals have produced the series of sand bars along the channel axes in the central basin shown in Figure 5 and in the western end of the east basin, as shown in Figure 6. Some of these shoals and sand bars have become vegetated with eel grass beds, as shown in Figure 6. The coarser-grained sediment infilling has also been a by-product of the succession of high energy El Nino winters that have occurred since the late 1970's. The same rains that produced runoff form Agua Hedionda Creek have produced even more significant runoff from the major rivers up coast, notably the Santa Margarita River and the San Luis Rey River. Major flooding events in these rivers since 1978 have produced large sandy delta deposits at the coast that have been transported southward by the prevailing littoral drift, as shown in Figure 9. As this littoral drift passes by the inlet to Agua Hedionda, the tidal flushing and net inflow of water through the inlet due to plant operations causes some of the littoral sediments to be ingested by the lagoon and form the inlet bar of the west basin shown in Figure 10. It is this bar which threatens closure of the ocean inlet. Although it is periodically removed by a biannual maintenance dredging cycle, the bar has grown more rapidly and larger in extent during the later portions of recent dredge cycles. The amount of these sediments which enter via the ocean inlet to build this inlet bar is in direct proportion to their updrift supply. That supply has been increased not only by Santa Margarita River floods of the last 15 years, but also by episodic harbor dredging and beach nourishment activities at Oceanside. fifilitiiirifiritftifiii 30 25 - 20 - o ocr 15 - c 'ro 10 - 5 - 1920-1945 1946-1977 1978-1994 O fW&&tfWW£V&y''/'VMy&kW&''W7W^W//W&:y,'^ ,,.-:-'.- • :,-.— i ii [ 'i' r V'| i i i | i i i | 'i i i | i i i | i i i | i i i [ t i i | i 'ii [ i i'i [ i i i |'ii i | i i i |'n 1920 1924 1928 1932 1936 1940 1944 1948 1952 1956 1960 1964 1968 1972 1976 1980 1984 1988 1992 1996 Rainfall time history for San Diego, California from Elwany, et al (in press). FIGURES: FIGURE 9: Prevailing littoral drift (time-averaged net longshore transport) along the Oceanside Littoral Cell. FIGURE 10: Inlet bar iniilling the recharge zone in the inlet channel section of the west basin, as viewed looking north. m 15 Coarser-grained sediments entering from the ocean inlet have been augmented by ^^^F H additional shoaling from local sources. Sandy shoals have formed along the north bank of the east basin due to wave erosion of the bank, and along the south bank due to slumping and bluffl^PI It erosion of the original dredge cut as a result of weathering by wind and rain. These bank shoals m are apparent in Figure 4, and some have also re-vegetated with eel grass. w «« H. TECHNICAL APPROACH •» To assess the effects of plant demand for cooling water and sediment infilling on inlet «M closure risk, a numerical modeling study was performed on a computer. The numerical *" modeling is divided into four major steps: <*» ** A) Tidal hydraulics and plant inflow modeling of the shoaled lagoon system to ** determine the tidal prism and scouring capability of the inlet (Section JH), *" B) Wave refraction and diffraction modeling to derive the delivery rates of sand to *"* the inlet by longshore transport (Section IV), C) Analysis of the risk of inlet closures (Section V), *"* D) Analysis of remedial measures to reduce the risk of inlet closures (Section VI). «* m The numerical tidal hydraulics analysis is performed, using actual inflow rates for a (*surrogate calendar year, from July 27, 1993 through July 27, 1994. Daily wave records from ** the nearby pressure sensor array off Oceanside, California, and tide gauge records from ScrippsDM Pier were also used over the same calendar year. This calendar year is assumed to be representative of typical user demand schedules, wave climate and rainfall. Because of the availability of these unbroken time series, it was possible to perform a risk _ analysis which calculates directly the temporal variability of the probability density and distribution functions for inlet closures over a maintenance dredging cycle. This procedure f» m yields closure probabilities day by day, allowing for identification of combinations of user demand, tidal epoch, and wave climate which contribute to enhanced or diminished likelihood m of inlet closure, as well as quantification of that likelihood. 4* Hi „ 16 The quantification of present levels of risk of inlet closures provides a baseline for w assessing the effectiveness of possible remedial measures. Three distinct classes of remedial measures are evaluated: |H ^ 1) Manipulation of plant inflow rates, <M 2) Modifications to the geometries of choke points within the lagoon, and «« 3) Dredging of the complete lagoon system. «• «•* Each of these remedial measures are evaluated for effectiveness by repeating the tidal hydraulics *"• and inlet closure risk analyses using forcing function data for the surrogate year of July 27, 1993 ««» through July 27, 1994. Having selected the most effective of these remedial measures by this ** procedure, a detailed set of tidal-hydraulic elevations and habitat acreage calculations are *» performed to determine environmental benefits. The habitat-benefits analysis utilizes predicted ** tides out to the year 2020 for extreme spring, neap and mean tidal ranges. flM ** JH. TIDAL HYDRAULICS AND PLANT IN-FLOW MODELING ** This phase of the analysis involves two sets of equations: 1) tidal hydraulics, which i^H computes the variations in water elevation and velocity throughout the lagoon for a variety of ** ocean tides driving the system at the ocean inlet, and 2) plant inflow dynamics which remove tar water from the lagoon system in a time-varying manner. These two sets of equations are solved ^ repeatedly for the existing lagoon bathymetry and plant flow rates, and for the various remedial*•options. DPI It A) Tidal Hydraulics Model The tidal hydraulics equations are solved on a Cartesian computational grid as configured in Figure 1. The grid boundaries are determined by the highest possible waterline m that could result from the maximum astronomic tide that will occur through the year 2020 (see Section UJ-B). The extreme higher high waterline of such a tide could be no higher than +4.26 IP m ft NGVD. The area within the +4.26 ft NGVD contour has been divided into 25 x 25 ft grid cells, as shown in Figure 1. The x-axis of this gridding system is directed in and out of the 17 lagoon; the y-axis is directed across the width of the lagoon; and the z-axis is directed vertically upward. The origin of the grid system is located at the center of the ocean inlet at California coordinate, x = 1,665,560 (ft easting) and y = 357,948 (ft northing). The elevation for z = 0 is at 0 ft NGVD. A depth, h, is assigned to the center point of each grid cell in Figure 1 according to June 1993 soundings represented in Figure 4, with updated local corrections in the center and east basins per November 1995 soundings. The depth integrated momentum equations which are solved for each grid cell in Figure 1 are: a, _ at °3y p^W w» *» where: g = the acceleration of gravity, ** f = the Coriolis parameter, *» h = the depth of the lagoon at grid cell center points relative to 0 ft NGVD, *" U, i? = the mean velocity in the x-, y-direction, ** i} = the deviation of the free surface, about z = 0 ft NGVD, k = the bottom friction coefficient. The water elevation at each grid cell, 17, is given by the continuity equation with a fluid sink " equivalent to the plant flow rate, Q, Ml 2 d*l + d(hU) + a (hi?) _ Q n\at ax ay o, + hp)b m m where b is the width of the plant infall channel, and hp is the depth of the plant infall channel ^ relative to z = 0 ft NGVD. tt 18 The finite-difference grid used was a standard Richardson lattice, with x- and y- directed velocities located along the sides of each grid and elevations and depths located in the center. The final finite-difference form of the equations of motion used are adapted from Sucy, Pearce, and Panchang (1989). For any grid cell x = i and y = j, the solution to equations 1-3 is given at time, t + 1, from solutions at the preceding time step, t, by: + t (4) Ax A + - ^, (5)Ax where A t is the time step length and sjtj = 0.25 * (r?!(j + 0<i(j+1 + i?Vi(j + t?Vi,j+i) (6) The < h > appearing in equations 4 and 5 is given by: There are analogous forms for the Y-directed momentum equation. Each successive value in time of T;, U, and t> is calculated in terms of previously calculated quantities making the numerical scheme explicit. The tidal hydraulics solutions given by equations 4 and 5 are driven by two forcing functions: 1) the tidal elevation of the ocean tides at the ocean inlet, TJO>O, and 2) the plant flow rate, Q. 19 The velocities calculated from equation (5) can be used to evaluate the threshold conditions for sediment transport at the inlet. In order to initiate sediment motion, the inlet velocity U0, must exceed: U0 ;> gD St 3/5 Re1'51 (8) 1 1/2 P St = ut/coD RC = Ul = Equation (8) specifies the threshold velocity, which is dependent upon the median grain size of the sediment, D. A grain size analysis was conducted by Leighton and Associates (1988) on the sediments which enter the ocean inlet at Agua Hedionda and form the inlet bar in the inlet channel shown in Figure 4. The results of their analysis is shown in Table 1. B) Ocean Tidal Forcing Functions There are two distinct sets of ocean tidal forcing functions used to drive the tidal hydraulics model. The first of these are the measured ocean tides for the surrogate year of July 27, 1993 to July 27, 1994 that are used in the inlet closure risk analysis. The second are predicted extreme tides used to assess the possible range of water levels and habitat acreages of the most effective remedial option for reducing inlet closure risks. For the risk analysis the measured tides from nearby Scripps Pier are taken from July 27, 1993 to July 27, 1994. In an earlier study, Jenkins, Skelly & Wasyl (1989), pressure sensors and current meters were placed in ocean waters off the lagoon inlet and the plant cooling water outfall. These measurements verified that the tidal amplitudes at Agua Hedionda are identical to those at Scripps Pier. The only measurable difference between the two tidal records is a 3.6 minute phase lead for the Agua Hedionda ocean tides. This is less than one time-step interval used in the hydraulic model, wherein each time step, At, is 6 minutes. Nonetheless, the Scripps Pier tidal record for the period July 27, 1993 to July 27, 1994 was shifted forward 3.6 minutes and sampled at 6 minute intervals to structure a tidal 20 TABLE 1 GRAINSIZE DISTRIBUTION OF AGUA HEDIONDA SEDIMENTS Microns 2000.000 1414.213 1000.000 707.107 500.000 353.553 250.000 176.777 123.000 83.388 62.550 44.194 31.250 22.097 pm -1.000 -0.500 0.000 0.500 1.000 1.500 2.000 2.500 3.000 3.500 4.000 4.500 5.000 5.500 Percent 0.000 0.016 0.000 0.063 0.973 9.616 37.835 39.953 10.212 12.302 0.031 0.000 0.000 0.000 Cum Percent 0.000 0.016 0.016 0.078 1.051 10.667 48.502 88.455 98.667 99.969 100.000 100.000 100.000 100.000 forcing file for the lagoon inlet. Examples from this file are plotted in Figures 11-13. These examples show both the extreme and average ranges in the spring-neap cycle for the surrogate calendar year. The largest diurnal tidal range for spring tides occurs twice annually during the perigean spring tides of February, shown in Figure 11, and in August, shown in Figure 12. The deepest neap tides with the smallest diurnal range also occur during these same perigean spring- neap cycles. The months in between the winter and summer perigean spring tides have spring neap cycles more typical of the annual mean diurnal tidal range, like that shown in Figure 13 for the month of April. ti n ri ri t ..... i t ..... i i i 1.5 1 - 0.5 - 1 • -0.5 -1 •1.5 SPRING - NEAP CYCLE: FEBRUARY, 1994 SCRIPPS PIER TIDE GAUGE, LA JOLLA, CA 8 12 16 20 TIME (DAYS) 24 28 FIGURE 11: Tide gauge record for Scripps Pier, California, February 1994 n f i t i ri ri ri n ri n SPRING - NEAP CYCLE: AUGUST 1993 SCRIPPS PIER TIDE GAUGE, LA JOLLA, CA CO I 10 15 20 TEME (DAYS) 25 30 FIGURE 12: Tide gauge record for Scripps Pier, California, August 1993 ri ri ri ti ti ri t i ri ti ti ri ri ri ti I SPRING - NEAP CYCLE: APRIL 1994 SCRIPPS PIER TIDE GAUGE, LA JOLLA, CA I I 10 15 20 TIME (DAYS) 25 30 FIGURE 13: Tide gauge record for Scripps Pier, California, April 1994 24 For the extreme tidal range habitat analysis of the best remedial option, the tidal hydraulics modeling effort uses a numerical forecast of the ocean tides at the lagoon inlet based upon the earlier work of Flick and Badan-Dangon (1989), Flick and Cayan (1984), and Flick (1991). For this purpose, tidal harmonic constituents derived from the nearby Scripps Pier tidal gauge were assimilated in a Fourier reconstruction of the predicted tides to the year 2020 AD. These predicted tides were assembled in 0.1 hour timesteps and corrected to NGVD (National Geodetic Vertical Datum). Mean sea level at the project site is +0.19 ft NGVD. The predicted tide records were then searched numerically for the tidal episodes having the maximum and minimum diurnal ranges. A maximum in the diurnal range occurs every four and one-half years, and the largest such tidal episode until the year 2020 is plotted in Figure 14. As seen on inspection of Figure 14, the highest HHW which the lagoon will ever experience reaches an elevation of +4.26 ft NGVD. The deepest low water level is found in Figure 14 to occur at an elevation of -4.58 ft NGVD. Thus the most extreme diurnal tidal range is found to be 8.84 ft. The minimum diurnal tidal range to be found among the predicted tides until the year 2020 is plotted in Figure 15. This tide will provide the minimal tidal forcing which the lagoon system will experience and will thus represent the condition of greatest susceptibility to inlet closure. We find, upon inspection of Figure 15, that the highest water level of this extreme neap tide reaches only +1.22 ft NGVD, while the LLW level reaches -1.89 ft NGVD. Thus, this extreme neap tide produces a diurnal tidal range of only 3.11 ft. While this condition is a time of enhanced vulnerability to inlet closure, such extreme neap tides occur only once every 18.6 years. While the extreme tide is used to bracket the tidal prism variability, the sustainable water levels and average habitat acreages are a function of the mean tides. For this assessment, the predicted tides until the year 2020 were analyzed spectrally to determine the mean diurnal range between MHHW and MLLW. The 30-year data block of predicted tides were then searched for the tidal month whose mean diurnal range most closely matched the mean diurnal range of the entire data block. That tidal month was determined to occur in October 1997 and is plotted in Figure 16. The mean diurnal range of the tides plotted in Figure 16 was determined to be 5.51 t i ii t i it r i i i r i t i r i r i t i i i t i • • i i i i i i EXTREME SPRING TIDE FOR THE PERIOD 1990-2020 PH Oi—i CO Q > O Oh—I E- 0 24 TIME ( HOURS ) FIGURE 14: Extreme Spring Tide for the Period 1990 - 2020 r i t t i i i i f i f i i i i i i i i f i i i i i i i t i i i f i EXTREME NEAP TIDE FOR THE PERIOD 1990 - 2020 Du O 00 Q > O o 5 4 - 3 -- 2 -- 1 / 0 -- _ i -2 -- — 3 -4 0 24 TIME ( HOURS ) 48 FIGURE IS: Extreme Neap Tide for the Period 1990-2020 f i r t f i t i t t f i f i i i f i f i i i i f i i i t i t i OCTOBER 1997 TIDE RECORD FOR CALCULATING TIDAL HYDROLOGY o > 168 504 672336 TIME ( HOURS RELATIVE TO 00:00 1 OCTOBER 1997 ) FIGURE 16: October 1997 Tide Record for Calculating Mean Tidal Hydrology 28 ft from MLLW to MHHW. This is identically the mean diurnal range to three significant figures for the 30-year block of predicted tides until the year 2020. C) Plant Flow Rate Forcing Function Plant flow rate data was provided in hourly increments on an ASCn-formatted diskette by SDG&E for the calendar year of July 27, 1993 to July 27, 1994. These data are plotted in Figure 2. Inspection of Figure 2 indicates that constant flow rates of 635 mgd are fairly representative of the high user demand months of mid-July to mid-September. Figure 17 gives a histogram of the plant flow rate file which indicates that this high user demand rate is the second most common level of activity, and that the most common level is between 555 and 585 mgd or about 70% of maximum capacity. It is also noted that inflow rates vary at most on a daily rather than hourly basis, and often remain constant for a week or longer. This is a fortunate circumstance from the hydraulic modeling standpoint because the tidal hydraulics model must be run at 0.1 hr time steps to insure numerical stability. The slow variability of the inflow rate data files allows interpolation to 0.1 hr increments without introducing aliasing errors. The plant inflow rate file was interpolated to 0.1 hr time step intervals, and input to the the tidal hydraulics model as a time varying sink (suction) positioned in the computational grid at the point labeled "Plant Infall" in Figure 1. D) Results for Existing Lagoon Bathymetry The tidal hydraulics model was run on the existing lagoon bathymetry using the ocean tides and plant flow rate forcing functions to determine the extreme and mean ranges of variability in the lagoon waterlines, water elevations, tidal prism and inlet velocities. The highest possible waterline with the largest extent of open water which the lagoon can presently achieve is shown in Figure 18. This waterline is the result of the higher high water level (HHW) of the most extreme spring tide to occur between 1990 and 2020 as shown in Figure 14. The corresponding lower low waterline (LLW) for this most extreme spring tide appears in Figure 19. Figure 19 shows the configuration of the present lagoon system when it has the least amount of open water, and clearly show the sand bars, shoals, and mud flats depicted in the photographs of Figures 5, 6, 7, and 10. These waterline simulations were performed for a •n mm ri ri n ri ri f i 11 i i r i ri r i r*iri ri SDG&E ENCINA POWER PLANT FLOW RATE HISTOGRAM JULY 27, 1993 • JULY 27, 1994 C/5 120 2 p 100 I 80 60 o1 40 20 0 ACTUAL PLANT FLOW RATE PL ANT OUTFALL FLOW RATE (mgd) FIGURE 17: Histogram of plant flow rates for the calendar year of July 27,1993 to July 27,1994. r i r i ii ii ii ii 11 11 ii ii Zm = +2.83 ft NGVD Zm = +2.21 ft NGVD HYDRODYNAMIC TIDE SIMULATION EXTREME SPRING TIDE HHW TOTAL WETTED AREA = 12,183,736 ft2 TIDAL PRISM = 63,636,515 ft3 INTERTIDAL AREA = 90.11 ACRES 0 500 1000 1500 8000 3500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.46 ft HEAD LOSS = 0.97 ft HEAD LOSS = 0.62 ft Zm = +3.29 ft NGVD DR SCOTT A. JENKINS CONSULTING AGUA HEDIONDA 570 mgd PLANT INFLOW SCOTT A. JENKINS PhD & JOSEPH WASYL FIGURE 18: Lagoon waterline for existing bathymetry at the HHW phase of the extreme spring tide for a plant flow rate of 570 mgd. r i 11 ti 11 ti ii 11 11 11 11 ii i f f • j = -4.19 ft NGVD = -3.55 ft NGVD HEAD LOSS = 0.39 ft HEAD LOSS = 0.64 ft HYDRODYNAMIC TIDE SIMULATION EXTREME SPRING TIDE LLW TOTAL WETTED AREA = 8,258,240 ft2 TIDAL PRISM = 63,636,515 ft3 SUB TIDAL AREA = 189.58 ACRES 500 1000 1500 ZOOO 2500 HORIZONTAL SCALE IN FEET j = -3.55 ft NGVD = -4.58 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA 570 mgd PLANT INFLOW SCOTT A. JENKINS PhD & JOSEPH WASYL FIGURE 19: Lagoon waterline for existing bathymetry at the LLW phase of the extreme spring tide for a plant flow rate of 570 mgd. 32 constant plant flow rate of 570 mgd, or 70% of maximum capacity, which is the most frequently observed level of plant activity according to Figure 17. The tidal prism associated with the volumetric change in lagoon water from the HHW waterline in Figure 18 to the LLW waterline in Figure 19 is 63,636,515 ft3, as compared to the original spring tidal prism of 80,000,000 ft3 just after completion of construction in 1955. Thus, there has been a 20.45% reduction in the diurnal tidal prism during maximum spring tides when the plant operates continuously at 70% of maximum capacity. This tidal prism loss is due to a combination of sediment infilling and plant demand. Nonetheless, the tidal prism exchange with ocean water at a maximum spring tide that was achieved in the existing lagoon did involve 90.1 intertidal acres, with a subtidal acreage of 189.6 acres. The shoals and bars apparent in Figure 19 have quite substantially increased the intertidal area and diminished the subtidal area relative to the original construction profile. According to FJlis (1954), the spring intertidal area of the original lagoon was only 58.1 acres, while the spring subtidal area was 235 acres. Inspection of Figures 18 and 19 reveals that the three basins which make up the lagoon system are at three different water levels during both the HHW and LLW phases of the spring tide. This is due to the head losses occurring at three separate constrictions (choke points) in the flow network. These choke points are: 1) the inlet channel with its 90° turn and inlet bar constriction (see Figures 10 and 19); 2) the railroad bridge; and 3) the 1-5 bridge with the associated sand bars of the central and east basin depicted in Figures 5 and 6. The degree to which the associated sand bars have exacerbated the constrictions in these choke points is especially apparent in the LLW waterline for spring tide shown in Figure 19. Because the dissipative work done by friction is greatly enhanced as the flow is accelerated through these choke points, the water level on the downstream side of each choke point is always less than the upstream side, regardless of whether the flow is ebbing or flooding. This action retards the ability of water to enter a basin on flooding tide or leave a basin on ebbing tide. For example, the 0.97 ft of head loss in the inlet channel and around the inlet bar prevents the west basin from ever reaching the +4.26 ft NGVD water level which the ocean reaches at HHW during the extreme spring tide. The accumulation of additional head losses at the railroad bridge and 1-5 bridge prevents the east basin from ever getting above +2.21 ft NGVD at the HHW level of the maximum spring tide. Similarly, water is constrained from evacuating the east basin by head 33 losses in the 1-5 bridge and railroad bridge choke points. As a result, the water level in the east basin never drops below -3.55 ft NGVD at the LLW stage of the maximum spring tide, despite the fact the LLW in the ocean reaches -4.58 ft NGVD. The continuous variation of the east basin water level throughout several tidal oscillations during the maximum spring tide is shown in Figure 20 for a plant flow rate of 570 mgd. The diminished tidal range of the east basin relative to the ocean tides reflects the dynamic losses of tidal prism through the effects of frictional damping primarily in the choke points. It is noted that there is less discrepancy between basin tides and ocean tides at low tide levels than at high tide levels. This is due to the suction action of the plant which aids in withdrawing water from the lagoon system during ebb tide and inhibits the ability of the lagoon system to completely fill during flood tide. It is also noted that the low tide levels are achieved later in the east basin relative to those in the ocean (phase lag), and that the low tide water level oscillations are somewhat distorted (second harmonic distortion). These are the result of friction and non-linear inertia! effects in the lagoon hydraulic response which are most pronounced in the shallow water configuration around the LLW tide levels. The basic features found in the spring tide response of the lagoon system are also repeated to a lesser degree in the response to forcing by mean tides at plant flow rates of 570 mgd (70% maximum activity). Figure 21 shows the lagoon waterline at mean higher high water (MHHW), while Figure 22 shows the waterline at mean lower low water (MLLW). The tidal prism exchange involved when the lagoon oscillates between its MHHW waterline and the MLLW waterline is 35,541,937 ft3, as compared to an original mean tidal prism of 55,000,000 ft3 immediately after the construction dredging in 1954. Hence the combined action of nominal plant demand for cooling water and sediment infilling has reduced the mean tidal prism by 35.4%. This tidal prism loss manifests itself in the tidal damping of the east basin mean tides plotted in Figure 23 relative to the mean ocean tides. The accumulation of head losses in the three choke points has muted the MHHW level in the east basin to only +1.36 ft NGVD as compared to an MHHW in the ocean of +3.16 ft NGVD. Plant suction augments the drawdown of water, such that MLLW levels in the east basin more closely approach those in the ocean, reaching -1.52 ft NGVD in the east basin as compared to -2.35 ft NGVD in the ocean. Because of the smaller tidal range for the mean tides relative to the extreme spring tides, the r I 11 11 • i II ii tt ii t i ii fi ii ii ii 11 it t i fi ii MAXIMUM SPRING TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN EXISTING CONDITIONS; AGUA HEDIONDA LAGOON, CA 5 Q > O O OCEAN TIDE EAST BASIN 20 26 TIME ( hrs ) 38 44 FIGURE 20: Water level variations for existing bathymetry in the east basin vs. the ocean throughout the extreme spring tidal cycle at a plant flow rate of 570 mgd. v i t i til* i i it 11 ri ri i i 11 t i t > 11 ti t i r i ti Zm = -1-1.92 ft NGVD HEAD LOSS = 0.83 ft HYDRODYNAMIC TIDE SIMULATION MEAN TIDE HHW TOTAL WETTED AREA = 11,530,775 ft: TIDAL PRISM = 35,541,937 ft3 INTERTIDAL AREA = 31.6 ACRES 0 500 1000 1500 2000 8500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.41 ft Z™ = +1.36 ft NGVD HEAD LOSS = 0.56 ft Zm = +1.36 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA 570 mgd PLANT INFLOW SCOTT A. JENKINS PhD & JOSEPH WASYL FIGURE 21: Lagoon waterline for existing bathymetry at the HHW phase of a mean tide for a plant flow rate of 570 mgd. f f i c i r i t i f • i ri t t t i i i • ifitttiiiriiifftiiri j = -2.04 ft NGVD HYDRODYNAMIC TIDE SIMULATION MEAN TIDE LLW TOTAL WETTED AREA = 10,154,644 ft2 TIDAL PRISM = 35,541,937 ft3 SUB TIDAL AREA = 233 ACRES 0 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.31 ft i = -1.52 ft NGVD HEAD LOSS = 0.52 ft j = -1.52 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA 570 mgd PLANT INFLOW SCOTT A. JENKINS PhD & JOSEPH WASYL FIGURE 22: Lagoon waterline for existing bathymetry at the LLW phase of a mean tide for a plant flow rate of 570 mgd. • i • I t 1 f t f i • i I t f i t * r i f i f * I 1 I i ff i f i t i • I t i MEAN TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN EXISTING CONDITIONS; AGUA HEDIONDA LAGOON, CA Q> O iz;o OCEAN TIDE EAST BASIN 14 20 26 32 TIME ( hrs ) 38 44 FIGURE 23: Water level variations for existing bathymetry in the east basin vs. the ocean throughout a mean tidal cycle at a plant flow rate of 570 mgd. 38 frictional phase lag and non-linear inertial distortions near low water are less apparent in Figure 23 (although still present) than those found in Figure 20 for the spring tides. The water levels throughout the east basin are essentially flat with most of the tidal damping, phase shifts, and second harmonic distortion occurring across the choke points. Inspection of Figures 21 and 22 indicates that tidal damping effects are progressively smaller for the central and west basins. Altogether, the lagoon in its present bathymetric configuration at 570 mgd produces a mean intertidal area of 31.6 acres and a mean subtidal area of 233 acres. The shoals and bars have slightly increased the mean intertidal area over the original construction profile, which had a mean intertidal area of 29.6 acres (see Ellis, 1954). These same shoals and bars have diminished the mean subtidal area rather substantially, which used to account for 251 acres in the original construction profile (Ellis, 1954). Figures 24 and 25 show the HHW and LLW waterlines due to the action of the extreme neap tides shown in Figure 15 when the power plant is operating at 570 mgd (70% of maximum activity). Because of the small tidal range during these neap tides, there is not a great deal of excursion in the waterline between the HHW phase of the tide and the LLW. Consequently, the tidal prism is only 18,022,439 ft3, which is not enough to meet plant demand at 70% of maximum activity. Plant flow rates at 570 mgd will consume 19,709,922 ft3 during any given ebb tide interval. Consequently, ebb flow will leave the lagoon system through the plant condensers, rather than the ocean inlet, causing the ocean inlet to become especially vulnerable to sediment infilling during ebbing neap tides. The small waterline excursions between HHW in Figure 24 and LLW in Figure 25 also contribute to a greatly diminished intertidal area totalling only 19.5 acres during neap tide. Because the LLW in Figure 25 is only slightly more than a foot below mean sea level in the east basin, the subtidal area during these neap tides remains relatively large, or 234 acres. The small tidal range of the extreme neap tide results in greatly diminished tidal currents through the choke points with correspondingly smaller head losses. Figure 26 shows that the accumulation of head losses during neap tide results in less tidal damping in an absolute sense relative to the spring and mean tide cases shown in Figures 20 and 23. The plant suction at 570 mgd substantially depresses the neap tidal oscillation in the east basin below the ocean tide. Because of the reduced choke point velocities, the phase lag and second harmonic distortion ! i i r v 11 i i r i t i i i i § i i t I f i ii ti 11 ii ii Zm = +0.49 ft NGVD HEAD LOSS = 0.51 ft HYDRODYNAMIC TIDE SIMULATION EXTREME NEAP TIDE HHW TOTAL WETTED AREA = 11,052,077 ft TIDAL PRISM = 18,022,439 ft3 INTERTIDAL AREA = 19.5 ACRES 0 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.22 ft Zm = +0.18 ft NGVD HEAD LOSS = 0.31 ft Zm = +0.71 ft NGVD Zm = +0.18 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA 570 mgd PLANT INFLOW SCOTT A. JENKINS PhD & JOSEPH WASYL FIGURE 24: Lagoon waterline for existing bathymetry at the HHW phase of the extreme neap tide for a plant flow rate of 570 mgd. ii i i • i it fi it t t ft §t ii r i ii it it it i i = -1.64 ft NGVD HYDRODYNAMIC TIDE SIMULATION EXTREME NEAP TIDE LLW TOTAL WETTED AREA = 10,200,617 ft' TIDAL PRISM = 18,022,439 ft3 SUB TIDAL AREA = 234 ACRES 0 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET Z; = -1.26 ft NGVD HEAD LOSS = 0.38 ft Z; = -1.89 ft NGVD = -1.26 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA 570 mgd PLANT INFLOW SCOTT A. JENKINS PhD & JOSEPH WASYL FIGURE 25: Lagoon waterline for existing bathymetry at the LLW phase of the extreme neap tide for a plant flow rate of 570 mgd. ii t i t i ft ft ft it 11 r i r i r i ii t i ii i i ri ii ii ti NEAP TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN EXISTING CONDITIONS; AGUA HEDIONDA LAGOON, CA Q> O o OCEAN TIDE EAST BASIN 8 14 20 26 32 TIME ( hrs ) 44 FIGURE 26: Water level variations for existing bathymetry in the east basin vs. the ocean throughout the extreme neap tidal cycle at a plant flow rate of 570 mgd. 42 around low tide levels is only vaguely apparent in the neap tidal oscillation of the east basin. The results discussed in Figures 18-26 were all calculated at a constant flow rate of 570 mgd, which was the most commonly observed level of plant activity. However, inspection of Figure 17 reveals that there were a very significant number of days during the surrogate calendar year for which the plant operated at much higher flow rates. These higher flow rates will diminish the spring, mean and neap tidal prisms even more than is shown in Figures 18 and 19, 21 and 22, or 24 and 25. To determine additional tidal prism losses during actual operating conditions, the tidal hydraulics model was run continuously for the tides and plant flow rates that existed for the entire year between July 27, 1993 and My 27, 1994. The resulting range of variations in tidal prism was found to be that shown in Figure 27. This figure represents the tidal prism variation over a year in terms of an incremental probability density function. The incremental probability density function was constructed by counting how many of the n = 703 realizations of tidal prism from the hydraulics model runs for the surrogate year lie within successive incremental limits. Figure 27 gives a plot of the incremental probability density function fv AV for tidal prism increments of AV = 1,000,000 ft3. The probability of the tidal prism occurring between V0 and V0 + AV is: VAV (9)P(VO <V(T) <V0 + AV) = | fv(V)dV where fv(V) is the tidal prism probability function. It is apparent from Figure 27 that the tidal prism probability density function peaks at 26,000,000 ft3. Hence, the mean tidal prism for annual operating conditions is actually less than that calculated in Figures 21 and 22 at 70% plant activity. In fact, the mean tidal prism for annual operating conditions is 52.7 % smaller than the original post-construction value (see Ellis, 1954) before sedimentation and expanded plant generating capacity impacted the lagoon hydraulics. Furthermore, the maximum realizable tidal prism for spring tide during this surrogate year was only 54,000,000 ft3, a loss of 32.5 % in maximum diurnal prism. Even more serious for inlet closure concerns is the fact that neap tidal prisms for annual operations fell on some days to only 1-2 million cubic feet. I i c i t i t i f i t t t i f i r i f ! c i f i f i i i t i t t I i i i I i TIDAL PRISM PROBABILITY DENSITY FUNCTION DUE TO EXISTING BATHYMETRY: 27 JULY 1993 - 27 JULY 1994 6 5- E- d53<CQo 0 0 10 20 30 40 50 60 70 TIDAL PRISM ( MILLION CUBIC FEET ) 80 FIGURE 27: Incremental probability density function of the tidal prism for existing bathymetry, tides and plant fow rates for the period between July 27, 1993 and July 27, 1994. 44 The inlet velocities during a typical spring-neap cycle of the surrogate calendar year are shown in Figure 28. This figure represents the combined effects of a maximum tidal range and high plant demand for cooling water, with plant flow rates varying between 635 and 669 mgd. Superimposed on the oscillating inlet velocities in Figure 28 are the threshold velocities of the sediment during both ebb and flood as calculated from equation (8) using the grain size data from Table 1. It is readily apparent, upon inspection of Figure 28, that the high plant flow rates have greatly strengthened the flood flow velocities and have weakened the ebb flow velocities. Consequently, the flood flow velocities exceed the threshold velocity of the sediment on every tide cycle, while only a few ebb flow cycles exceed the sediment threshold velocity, and even then, only for a relatively brief period of time. Thus, high plant flow rates preferentially offset the inlet velocities such that sediment can be readily transported into the lagoon during flooding flow, but are rarely expelled from the lagoon by weakened ebb flow currents. This effect is greatly exacerbated during the neap tides, as shown in Figure 29. Here it is shown that there is virtually no ebbing flow out of the ocean inlet during a neap tide when plant demand for cooling water is at a moderately high level of 587.5 mgd. As was concluded previously from Figures 24 and 25, the plant will consume a greater volume of water during an ebb neap tide interval than the available tidal prism. Consequently, the ebb flow during neap tides leave the lagoon system through the plant condensers rather than through the ocean inlet. Flooding neap flows, however, are shown in Figure 29 to still exceed the threshold velocity of the sediment. Consequently, the inlet flow becomes a one-way transport pathway: sediment enters the inlet due to above-threshold flooding flow, but no sediment is scoured from the inlet channel in the absence of any ebbing flow. This condition is clearly a time when the inlet becomes vulnerable to closure. IV. WAVE REFRACTION/DIFFRACTION MODELING AND LONGSHORE TRANSPORT RATES The lagoon inlet cannot ingest sand during the closure process any faster than sand is made available to the inlet by the wave transport processes. These processes are driven by the local wave properties which are calculated from wave refraction and diffraction modeling. Both distant swell waves as well as locally derived wind waves can be active in the inlet closure *fi***s***™"i<" ii it • t t i r i r i ti c i f 1 r i ft 11 ii ft ii it ti i i I i INLET VELOCITIES, CHANNEL AXIS AT HYW 101 BRIDGE EXISTING CONDITIONS: SPRING - NEAP CYCLE o CD 4- 3 E-"\ — I O O X -2 + -3 -4 -5 EXISTING CONDITIONS THRESHOLD VELOCITY EBB V 320 340 360 380 400 420 440 460 480 TIME ( HOURS RELATIVE TO 00:00, 1 AUGUST 1993 ) FIGURE 28: Inlet velocities during a maximum range spring-neap cycle for existing bathymetry and actual plant flow rates. f i ft II I I I i ff 1 II f 1 II f 1 f 1 II II II II II II 11 II INLET VELOCITIES, CHANNEL AXIS AT HYW 101 BRIDGE EXISTING CONDITIONS: NEAP TIDES o 4 -- 3- O O -J X< E-*H Jjz; -1 - -2 -- -3 -4 -5 - EXISTING CONDITIONS ••- THRESHOLD VELOCITY FLOOD A EBB V 140 160 180 200 220 240 260 280 300 TIME ( HOURS RELATIVE TO 00:00, 1 AUGUST 1993 ) FIGURE 29: Inlet velocities during a minimum neap tide epoch for existing bathymetry and actual plant flow rates. 47 process depending upon the particular diurnal tidal range. This results from the fact that both the wave height and the wave direction are important in determining the magnitude of the wave- driven longshore transport rates. For instance, relatively small wind waves which approach the shoreline at a large oblique angle are capable of producing longshore transport rates comparable to those of large waves that approach the shoreline nearly normally incident. Figure 30 reveals that waves cannot arrive at Agua Hedionda Lagoon from distant sources at arbitrary directions, due to the sheltering or shadows cast on the mainland by the offshore channel islands. There are two predominant "wave windows" through which distant swell waves can reach the mouth of the Agua Hedionda Lagoon. The largest, most open wave window looks to the southwest, allowing waves to approach the site from as far west as the south bank of San Clemente Island from a direction of 245.5° true. The south window is bounded further to the south by the intervening land mass of Baja California, which closes off the window at 165° true. A smaller window is open to the northwest for waves approaching the site through the gap between San Clemente and Santa Catalina Islands. This is the west window which is open from 261° to 277.5° true with a small shadow in the center due to San Nicholas Island between 270° and 272.5° true. The only waves which can arrive at the mouth of the Agua Hedionda Lagoon from directions outside of the apertures of either the south or west window are local wind waves generated inside the channel islands. The most predominant local wind waves are those due to the northwesterly diurnal sea breeze approaching the site from 296° true to 310° true which define a north window. Local wind waves from southerly winds due to Catalina eddies are shadowed by Point La Jolla. A) Wave Forcing File The California Department of Boating and Waterways and the U.S. Army Corps of Engineers jointly support a wave monitoring program called "Coastal Data Information Program", or CDIP. This program maintains an array of four bottom-mounted pressure sensors located off Oceanside Beach, California, at North Latitude 33° 10.7', West Longitude 117° 23.6' in an average depth of water of 36 feet MSL. Wave heights, directions and periods from this array are averaged over 6 hour intervals and reported in tabular format in monthly summary FIGURE 30 Wave windows at Agua Hedionda due to island sheltering 120 49 reports. Data from the reports spanning the surrogate year from July 1993 through July 1994 (CDIP, 1993, 1994) were compiled and entered into a structured preliminary data file. The data in the preliminary file represent partially shoaled wave data specific to the local bathymetry around Oceanside beach (see NOAA, 1972, and NRS, 1984). To correct these data to Agua Hedionda, they are entered into the refraction/diffraction numerical code described below, back- refracted out into deep water, and subsequently brought onshore into the immediate neighborhood of Agua Hedionda. Hence, the deep water wave data off Oceanside may be used to hindcast the waves at Agua Hedionda. B) Refraction-Diffraction Code and Gridding Systems Oceanside CDIP wave data are shoaled into the beaches near the project site using a numerical refraction-diffraction computer code called OCEANRDS based upon the parabolic equations of Kirby (1986). These equations are lengthy and tedious, so a listing of the core portion of this numerical code is given in Appendix I. This algorithm accounts for both changes in the direction of wave propagation (refraction) as well as scattering of incoming wave energy due to complex surrounding topography such as intervening submarine canyons or variations in the width of the continental shelf. The computer code also accounts for complicated nearfield wave reflection and diffraction patterns arising from waves reflected by shelf or canyon slopes or other abrupt changes in the bottom slope. These near-field wave diffraction processes are superimposed on the local swells and contribute to longshore variations in both wave heights and wave directions, which ultimately regulate the wave-driven transport and littoral drift. To perform these complex computations, the refraction/diffraction algorithm requires fine-scale resolution of the bottom bathymetry. A detailed bathymetry grid was developed for the model in order to propagate distant swells and local wind waves towards the mouth of the Agua Hedionda Lagoon. Depth soundings from the National Ocean Survey Digital Data Base for a latitude range of 33.00° to 33.2° and longitudes of 117.26° to 117.44° were used to create a 3 x 3 second of arc bathymetry grid with 300 ft x 300 ft spatial resolution and a depth range of 0 to 1000 ft. Surveys in the NOS database extended from the 1930's to the 1980's. In areas where two or more surveys overlapped, the most recent survey was used. The completed grid was screened for bad points through a comparison with NOS nautical charts. 51 In order to account for any conceivable far-field effects which offshore bathymetry might have on local wave properties, the refraction-diffraction grid was extended from the Carlsbad Submarine Canyon to the north of Camp Pendleton. For computations of the littoral drift rates and cross-shore transport rates to be used in the inlet closure probability calculations later in Section V, a near-field refraction-diffraction computational grid was centered around the lagoon inlet at latitude 33.146°. The nearfield refraction-diffraction grid spanned 11.4 statute miles of shoreline and yielded calculations of the wave heights and wave angles at 300 foot intervals along this reach of shoreline. C) Refraction-Diffraction Results Energy density plots for two extreme example storm events are shown in Figures 31 and 32. Figure 31 plots the spatial distribution of wave heights from the largest waves through the south window due to the September 24, 1939 which generated the ultimate design wave used in coastal engineering in Southern California. The south angle of wave approach is clearly visible in the beam spreading and beam converging patterns of the wave energy density over the shelf and near the surfzone. The surfzone is also clearly visible by the dark blue tones of the contour plot from the breakpoint to the shore. This indicates the reduction in wave height after wave breaking. The longshore variations in wave heights are the result of local convergence and divergence due to the refraction and diffraction patterns induced by the complex offshore bathymetry. These variations in wave heights along the shore will produce variations in the local longshore transport rate. Similar variations are shown in Figure 32, which shows the largest swells to pass through the west window due to the storm of April 22, 1958. Here longshore variations in wave heights are much more apparent due to the longer periods of these waves yielding more local diffraction effects. However, the maximum breaker heights are somewhat less due to the smaller deep water swell heights of the 1958 storm relative to the 1939 storm. Whereas the 1939 storm produced breaker heights on the order of 8.5 - 8.9 meters, the April 22, 1958 storm produced local breaker heights ranging between 6 and 7 meters. It is also apparent from the local beam spreading and beam convergence that the wave direction of the 1958 storm was more obliquely I I I I I I I I I I CARLSBAD, CA SEPTEMBER 24, 1939 STORM EVENT 33.150- 03 33.100 33.050 H 117.30 Longitude 2.0 4.0 6.0 8.0 10.0 Wave Amplitude ( m ) FIGURE 31 Refraction/Diffraction pattern for the September 24,1939 storm I I I I I I I I CARLSBAD, CA APRIL 22, 1958 STORM EVENT 33.150 H tl 33.100eg 33.050 117.40 117.30 Longitude 2.0 4.0 6.0 8.0 10.0 Wave Amplitude ( m ) FIGURE 32 Refraction/Diffraction pattern for the April 22, 1958 storm I I I I I I I I I I I I I I I I I I I 54 incident than that of the 1939 storm leading to more complex diffraction patterns, particularly around Carlsbad Submarine Canyon. The CDIP wave data measured off Oceanside, California between July 27, 1993 and July 27, 1994, as described in Section IV-A were entered into the refraction/diffraction numerical code listed in Appendix I and back-refracted out into deep water. An example of a back- refracted result is shown in Figure 33. Inspection of Figure 33 shows that deep water conditions are quite uniform. Hence, the deep water wave data derived from back-refraction of the Oceanside measurements gives a reasonable estimate of the deep water wave statistics for the small inner wave grid used in local transport calculations for the neighborhood of Agua Hedionda. The deep water wave data files derived from back-refraction are used as the primary input file for a wave refraction-diffraction analysis over the local bathymetry offshore of Agua Hedionda with the gridding and methods detailed in Section IV-B. By this procedure, the island sheltering effects on the incident deep water waves are backed out of the Oceanside array measurements and then applied to the deep water boundary of the local Agua Hedionda refraction/diffraction grid. These back-refracted deep water wave files are plotted in Figure 34 for heights and periods, and in Figure 35 for heights and direction relative to the shoreline normal at Agua Hedionda. Inspection of these figures indicates that the largest waves occurred between February 8 and 10, 1994, at days #196-198 of the surrogate calendar year from 27 July 1993 to 27 July 1994. Although the deep water waves reach 261 cm (8.56 ft) in height, inspection of Figure 11 reveals that this storm happened to coincide with a perigean spring tide maximum. Therefore, the action of these large waves in increasing sediment delivery to the lagoon inlet should be offset somewhat by the large tidal range and tidal prism which will act to flush the lagoon inlet. A probability density function can be computed for breaker height increments AH about any given wave height, H0, according to: H0+AH P(Ho<Hb<Ho + AH) = | ^(H)dH,, (10) I I I I I I I I I Back-Refraction of Oceanside Array Data, February 8,1994 33.30 33.20 - 117,50 117.40 Longitude 01234 Wave Height (m): Period = 11 sec Direction from = 300.5 degrees FIGURE 33 Back-refracted wave data from the CDIP Oceanside pressure sensor array DIRECTIONAL WAVE DATA FROM OCEANSIDE ARRAY JULY 27, 1993 - JULY 27, 1994 300 250- 200- 2 150- BJ < 100- * WAVE HEIGHT (cm ) i n in inn mi in u i ii • ID ii ii :••• iiinnii i linn mi i i in HI • i II I II I; i WAVE PERIOD I r25 -20 II ! II I i • ii •! mi til II II 18 I III 15 Q II -10 g 0 0 50 100 150 200 250 300 350 7-27-93 TIME (DAYS) 7-27-94 FIGURE 34 Deep water wave height and period time history from back-refraction of the Oceanside array measurements DIRECTIONAL WAVE DATA FROM OCEANSIDE ARRAY JULY 27, 1993 - JULY 27, 1994 300-i -40 -40 0 50 100 150 200 250 300 350 7-27-93 TIME (DAYS) 7-27-94 FIGURE 35 Deep water wave height and direction time history from back-refraction of the Oceanside array measurements ri ii ri ft ti ti ri t i t i ti ri ti 11 ri 11 11 ri ft WAVE HEIGHT PROBABILITY DENSITY FUNCTION 27 JULY 1993 - 27 JULY 1994 0 50 75 100 125 150 175 WAVE HEIGHT (cm) 200 225 FIGURE 36 Breaker height incremental probability density function evaluated at Agua Hedionda tidal inlet ri rl V1 WAVE ANGLE PROBABILITY DENSITY FUNCTION 27 JULY 1993 - 27 JULY 1994 *K5 3 1 o — 7- 6- 5- 4- 3i 1 -j n "\J 230 240 250 260 270 280 290 300 WAVE ANGLE (deg) 310 \ \ 320 330 FIGURE 37 Breaker angle incremental probability density function evaluated at Agua Hedionda tidal inlet (direction waves from relative to true north) 60 The incremental breaker height probability density function, fHAH, is calculated from the results of the refraction/diffraction analysis evaluated at the lagoon inlet for a wave height increment AH = 5 cm as shown in Figure 36. Comparison of the maximum observed wave heights to the extreme event waves in Figures 31 and 32 indicates that the waves of the surrogate year as represented in Figure 36 were rather average in height and devoid of any extreme events. Thus, the closure risk evaluated on the basis of these wave statistics in the following sections will be representative of average interannual fluctuations and will not reflect the high-risk conditions of episodic extreme event storms. The longshore transport acting to close the inlet is dependent not only on wave height but also on breaker angle, ab, at the lagoon inlet. The probability density function for a wave direction increment Aa about a given breaker angle, a0, may be defined from P(a0 < ab < a0 + Aa) = f»dab (11) The incremental probability density function for the breaker angle at the lagoon inlet, f«AQ;, is plotted in Figure 37 using a direction increment of a = 5°. Breaker angles are expressed as the direction the waves are from relative to true north. The shoreline normal at Agua Hedionda faces 242° true. Hence the preponderance of waves during the surrogate year were refracted toward the local shoreline with a wave angle directed from the northwest. Consequently, the most persistent direction of longshore transport throughout this period was southward in the neighborhood of Agua Hedionda, as diagrammed in Figure 9 for the general long-term littoral drift. D) Longshore Transport Rates and Updrift Dredge Disposal The refraction-diffraction calculations are used to estimate the delivery rates of sand to the inlet from computations of littoral drift estimated by the potential longshore transport rates. The formulation for the longshore transport rate is taken from the work of Inman & Bagnold (1963) according to: 61 I, = K [Cn • Syjb (12) where I, is the longshore transport rate; Cn is the phase velocity of the waves; Syx = E sina cosa is the radiation stress component; E = l/8pgH2 b is the wave energy density; p is the density of water; Hb is the near-breaking wave height; and K is the transport efficiency equal to: K = 2.2 c (13) crb where: /3 is the beach slope and co is the radian frequency = 27T/T, and T is the wave period. These equations relate longshore transport rate to the longshore flux of energy, which is proportional to the square of the near-breaking wave height and breaker angle. By this formulation, the computer code calculates a local longshore transport rate for each grid point in the refraction-diffraction calculation at 300 foot intervals along 11.4 miles of shoreline. After inputting the files of breaker heights, angles and phase speeds from the refraction analyses into equations (12) to (14), the variation in potential longshore transport along the local shoreline is calculated as shown in Figure 38. Inspection of Figure 38 indicates that if disposal is required north of the inlet, the best site with minimum southward drift over the widest variance in wave direction would be at decimal latitude 33.161°. This corresponds with the darkest section of the shadow zone identified in the refraction/diffraction plots in Figure 32. The diminished wave heights of the shadow zone reduce the southward directed potential longshore transport rate, making it the best suited location for placing the dredge material. Unfortunately, this region is 1,896 ft north of the permitted disposal site region. Closer inspection of Figure 38 reveals that the potential southward transport due to long period swells progressively increases from Oak Street south throughout the permitted disposal region. Only 62 the short period wind waves at the northernmost wave angles reach a transport minimum in the disposal area, and these transports are everywhere less than those of the long swells from the other wave angles. Therefore, the particular location within the permitted disposal area having the overall lowest southward drift is at Oak Street, as indicated in Figure 38. V. ANALYSIS OF THE RISK OF INLET CLOSURES The condition for hydraulic closure of the inlet is reached when a sufficient volume of sand has influxed to fill a region in the inlet channel section of the west basin known as the "recharge zone" (see Figure 4). The volume of sand required to fill the recharge zone is referred to as the "critical storage volume," Jcrit. For a typical post-dredging profile of the west basin, the critical storage volume below mean sea level is Jcrit = 6,349,073 ft3. The rate at which the recharge zone is filled is referred to as "the recharge rate" or R(t), and the volume of sand which influxes the recharge zone in some time interval, At, is referred to as the recharge volume, JR, written JR = R(t) dt (15) The critical condition at which the inlet is considered closed in a hydraulic sense occurs when sufficient time has lapsed, t = tR, for the recharge volume to equal the critical storage volume, or t0 +tR JR= J R(t)dt-J.rit (16) t0 A) Recharge Rate There are two opposing transport mechanisms involved in determining the recharge rate. The first of these mechanisms is longshore transport within the surf zone due to waves breaking at a non-normal angle of incidence to the shoreline. Because wave direction can vary throughout interannual and climatic cycles, the gross longshore transport must be f f II I » II II II II II II II II II II II II II II II I I c• 1—I 6 CO 6 O O.OQ O Ken O O LONGSHORE VARIATION OF LITTORAL DRIFT RATES: AGUA HEDIONDA, CA: DREDGE DISPOSAL SITE ANALYSIS 3.0 0.0 2.5- 2.0- 1.5- 1.0- 0.5 + /< LOWER WEST WINDOW, 261° MIDDLE WEST WINDOW, 270° UPPER WEST WINDOW, 277.5° LOWER NORTH WINDOW, 296° UPPER NORTH WINDOW, 310° K!E-iD!01 Oioj II IH \<iO 33.06 33.08 33.10 33.12 33.14 33.16 33.18 33.20 SHORELINE POSITION, DECIMAL DEGREES OF LATITUDE FIGURE 38 Longshore variations of southward directed potential longshore transport rates. 64 considered in this process. The inlet capture radius can intercept sand from either upcoast or downcoast longshore transport, and the gross longshore transport is often several times greater than the net longshore transport. For example, in the Oceanside littoral cell (see Figure 9), the net longshore transport of sand which determines the littoral drift has been calculated to average about 260,000 cubic yards per year. On the other hand, the gross longshore transport of sand due to short-term fluctuations in wave direction has been calculated at over 1 million cubic yards per year (see Inman and Jenkins, 1983). The gross immersed weight longshore transport, Igross, is the primary sand delivery mechanism to the recharge zone. It is opposed by tidal flushing by the ebbing flow whose immersed weight transport rate is 1^. Thus the recharge rate may be written R(t) = Igro" " Ifluih (17)(P, - P)gN0 where p is the density of seawater; p. is the density of the solid grains; g is the acceleration of gravity; and N0 is the volume concentration of the sedimentary deposition in the recharge zone. For well-sorted sand, N0 = 0.6, see Inman and Bagnold (1963). The immersed weight transport rates in Equation (17) are evaluated based upon the cohesionless granular transport mechanics of Bagnold (1963 and 1966). Accordingly, the transport is partitioned into bedload and suspended load to accommodate the motion of sediment within the water column and along the bed. The suspended load immersed weight transport per unit width, i,, is given by: . _ .D ' W - M|8 where c, is the suspended load transport efficiency, CD, is the drag coefficient, p is the density of the water, u is the velocity of the water, w is the settling velocity of the sediment, and & is the slope of the bed. Similarly, we write the bedload immersed weight transport per unit width, it,, according to: 65 ebCnpu3 i = L_Bl— (19)b tan<£ - u/3/|u| where tan 0 is the angle of internal friction of the sediment, and % is the transport efficiency for bedload. The water velocity appearing in the basic transport equations, u, will be represented in terms of a time-dependent wave velocity, U, which is perturbed by the inlet tidal velocity, U0, according to: u = U + U0 (20) Because the transport relations in equations (18) and (19) contain high powers of a time- dependent velocity and products with the modulus of the velocity, we invoke the following expansion under the assumption that the tidal velocity varies slowly relative to the wave velocity: u"\u\ = Un\U\ + (« + l)U0U"-l\U\ + Wn+ U0Un-2\U\ + ... (21) The oscillatory wave velocity, U, in equations (20) and (21) will be specified according to Stokes' second-order theory, which in the inner shoaling zone near the break point, may be written (see Flick, 1978): U = MJ cosof + «2cos (2<rt + 6) 2swbkh (22) sinh4#i Here, H is the wave height, co is the radian frequency, k is the wave number, B is the phase angle of the second harmonic, and h is the local depth of water. The parameters (H, a>, k) are determined by the refraction/diffraction analysis in Section IV-C, while the tidal inlet velocity is determined from the tidal hydraulics modeling in Section ffl-D. 66 The fraction of the gross longshore transport which contributes to recharge is, in large part, limited by the capture radius, Xr, of the jettied inlet system. The capture radius is controlled by the depth of scour at the mouth of the inlet jetties, Z,, and by the beach slope, as determined by the beach profile. The beach profile is determined from wave parameters using the theory of Keulegan and Krumbein (1949), and Inman, Elwany, and Jenkins (1993). Accordingly, the beach profile is specified as: h = Arf* ,3~\ 2/5 = ["0.68pCAVFg»3"| L 8m'w J »' = (P. ~ where CA and m are empirically derived factors, which fall within the ranges of 0.3 ^ m ^ 0.6, and 2.1 < CA <, 4.6, according to Inman, Elwany, and Jenkins (1993). The particular values for CA and m within these ranges is selected using calibration data of seasonal beach profiles. Hence, the capture radius, X,, will be written: fzl1""X, - \j\ (23) The recharge rate due to gross longshore transport is obtained by integrating numerically the longshore transport relations due to Komar and Inman (1970), or: . !_ AR {, E = UBpgtf 1.2U* <**> 1/2 b Hba 2kh 67 Here, ECn is the energy flux of the waves; E is the wave energy per unit area; Cn is the wave group velocity; v(x) is the longshore transport velocity; and ab is the wave breaker angle. In equation (24) all variables with a subscript "b" are evaluated at the breakpoint. The longshore transport velocity, v(x), is determined from the longshore current theories of Longuet-Higgins, (1970), according to: v(x) = v H JL -1 JLlnJL"149 X, 7 Xb Xb . v *2 JL ifx> X (25) ° 49 X where Xb is the width of the surf zone, or X* s 5/4 Hb tan/3. Inspection of equations 24 and 25 reveals that the longshore transport is strongest in the neighborhood of the breakpoint, x = Xb, where the longshore currents approach a maximum value of v (x) = v0. Consequently, if the capture radius of the inlet is small relative to the width of the surfzone, X, < Xb, then recharge due to gross longshore transport will be a relatively small fraction of the total surfzone transport. If, on the other hand, the tidal flow scours a deep bed profile at the inlet mouth, or if the waves are small such that the surfzone width is relatively narrow, then virtually all of the gross longshore transport will be intercepted and delivered to the recharge zone. Tidal flushing can occur simultaneously with recharge during the recharge period, and act to retard the rate of recharge, according to equation 17. We calculate the transport due to tidal flushing during recharge by averaging the bedload and suspended load transport rate per unit width over the period of ebb flow, T6, and by integrating across the width of the inlet mouth, Yr, according to: 68 T. Y = Y \ ' } r (i8 + 0 dydt (26) o o Because the recharge time is long in relation to the period of the waves, the time averages of terms in the power series expansion of the velocity, u which involve odd powers of n will vanish yielding the following volume flux due to tidal flushing: Ifhish Here, a threshold of motion criteria due to G.I. Taylor (1946) per equation (8) is applied to the tidal velocity in order for flushing to continue during the recharge period. If the tidal velocity drops below the value indicated in equation 8, then the transport from the capture radius due to tidal flushing will vanish, and the recharge zone will fill at the maximum recharge rate. The tidal velocity, U0, is in direct proportion to the tidal prism, whose variability throughout the surrogate year was found to obey the probability density function shown in Figure 27. B) Closure Probability Computations The tidal prism, breaker heights, and breaker angles vary with time throughout a calendar year; hence, we write: V =V(t) Hb = Hb (t) (28) <*b = <*b (t) There are n = 1405 points in 6.25 hour increments which make up the files for equation (28). These are numerically differentiated with respect to time to define the rate files: 69 A\7V =,, _ av Hb' = at 3H,b at at where primes denote time derivatives. Because of the time dependence of (V, H^ ab), the probability density functions (fv, FH, fj are also time-dependent and may be expressed: fv(t) = V'fv(V) fH(t) = H'fH(H) (29) fa(t) = a'fa(a) where fv(V), fH(H), and f«(a) are defined by equations (9), (10), and (11). The probability density functions through the calendar year of July 27, 1993 through July 27, 1994 have been determined for the tidal prism, fv(v), in Figure 27; for the breaker height, fH(H), in Figure 36; and for the breaker angle, ftt(a), in Figure 37. The inlet closure probability, Pc(t), may now be defined by the probability that the recharge volume at time, t, is greater than or equal to the critical storage volume, or PC (0 = P(JR(t) * Jen.) = P [ J Iflu* dt <£ J U. dt - Jcrit (p. - p) gNJ (30) Substituting equations (9), (10), (11), (24) and (27) into equation (30) per Boas (1966) yields a closure probability for any time, t = t0: 70 <t <t0 + At) f. (O dt to At q. MO^CO f fv(V)dVdt f. 0J o where lo+At o - - (3D o (•• = J p I § —l b 1 > and 2tan0 J and fc(t) is the closure probability density function. Accordingly, the inlet closure probability distribution function, which defines the accumulation of closure probability or accumulated risk of closure up to any given point in time, to, is Fc(t0) = J fH(t)f«(t) f fv (V)dVdt (32) C) Closure Probabilities for Existing Lagoon Bathymetry Plots of the closure probabilities, Pc(t0) = fc(to)At, are calculated for each month using equation (31), presented in graphic form in Appendix n. Closure probabilities are based upon a biannual dredge cycle under the assumption that the calendar year of 27 July 1993 to 27 71 July 1994 is a typical year (stationarity assumption) and can be folded over into a second year. If the dredging is performed annually, then closure probabilities would be approximately half as large as those presented. The months having the highest, or most sustained high, closure probabilities for the calendar year are shown in Figures 39-41. In these text figures, the closure probability appears as the blue line according to the vertical axis on the left-hand side. The actual or "normal" plant inflow rate recorded by SDG&E for each time increment is the red line per the scale read from the vertical axis on the right-hand side. In the black and white monthly plots of Appendix n, the closure probabilities are the solid line; the normal plant flow rates are shown by the dashed-dot-dot line. The highest closure probability found anywhere in the calendar year occurred in August 1993, in Figure 39, where Pc (t0) = 0.38 for t0 = 06Z August 22, 1993. Other months with sustained high closure probabilities were February, 1994, in Figure 40, and April, 1994, in Figure 41. In both February and April, a number of episodes where Pc - 0.26 were calculated. The high closure probability calculated in August is interesting from the standpoint that it did not occur during an episode of unusually high waves. Instead, it was the result of a combination of aggravating factors, including: 1) an oblique swell which refracted to make a large breaker angle, ab ~ 20° relative to the shoreline normal at Agua Hedionda, and 2) high sustained plant inflow rates between 600 and 635 mgd for the entire month. Although the diurnal tidal range was approximately within the mean range, the plant withdrew a sufficiently large volume of water away from the inlet during ebb tide that an insufficient amount of cross-shore directed tidal energy was available to compete with the large longshore flux of wave energy resulting from large breaker angles. On the other hand, the high closure probabilities in the months of February and April were the result of either large waves from winter storms or small diurnal tidal ranges during neap tides. Fortunately, these two conditions did not coincide. For example, the high closure probability episodes between February 16-20 in Figure 40 coincided with neap tides, while the high closure probabilities between February 22 and 27 were due to high waves while plant inflow rates maintained significant levels near 600 mgd. The largest waves of the entire calendar year, occurring February 8-10, coincided with the biggest spring tides, and consequently did not yield the highest closure probability for the year, although Pc - 0.26 during this period. In the month of April, the neap tides were only slightly smaller than the 72 mean tides, and hence, all the episodes of high closure probability were due to high waves from winter storms. The episodes of high closure probabilities calculated between February and April 1994 explain why maintenance dredging activities at that time in the west basin of Agua Hedionda had difficulty keeping up with the high rates at which sand was entering the lagoon. A histogram of the closure probabilities in Figure 42 summarizes the numbers of days in a two-year dredge cycle that elevated closure probabilities will occur. If an annual dredging cycle is employed, the values would be approximately half as large. The most noteworthy feature in Figure 42 is that enhanced closure episodes like those in Figures 39-41 where Pc ~ 0.26 occur 103 days in a two-year dredge cycle. Even more significant is that closure probabilities exceed 10% during 361 days out of the 730 days between maintenance dredging events. Interpretation of this figure, however, should consider the fact that the closure probabilities plotted in Figures 39-41 and in Appendix n represent incremental probability density functions, fc(t0)At, i.e., the rate of accumulation of probability over time increments of At = 6.25 hrs for the wave, tidal and inflow rate conditions which persist at any given instant in time, t = t0. Another relevant risk analysis statistic is the closure probability distribution function, which is calculated from equation (32) and shown in Figures 43 and 44. These plots represent the accumulation of risk over time. The solid line in Figures 43 and 44 gives the accumulation of closure probability for normal plant operations based upon the inflow rate time history in Figure 2. Figure 43 shows the accumulated risk of the first year of a two-year dredge cycle; while Figure 44 gives the accumulated risk at the end of a biannual dredge cycle. Note that after two years, the risk of inlet closure reaches a fairly significant 22 %. Under the assumption that the wave, tide and inflow rate statistics are stationary for the calendar year 27 July 1993 to 27 July 1994 (i.e., a typical year), the risk of inlet closure can be expected to reach 33 % for a three-year dredge cycle. Closure is more probable than not, Pc > 0.5 after 4.5 years. m to * i n t i r~i ri r i r i r i r i r i r i ri r i t i t i CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE AGUA HEDIONDA LAGOON , CA AUGUST 1993 0.5^ 0.4- • CLOSURE PROBABILITY NORMAL FLOW ADJUSTED FLOW 15 20 25 TIME (DAYS) r- 1000 - 800 -a -600 -400 200 30 FIGURE 39 Closure probability for the month of August during a biannual dredge cycle mm «*•ri r i ri n f i f ..... i ri CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE AGUA HEDIONDA LAGOON , CA FEBRUARY 1994 0.5^ 0.4- 8 0.3- 0.2- 0.1 - •CLOSURE PROBABILITY NORMAL FLOW A ADJUSTED FLOW A A t 8 12 16 20 TIME (DAYS) r- 1000 800 -g, 5 A A -600 400 -200 B 24 28 FIGURE 40 Closure probability for the month of February during a biannual dredge cycle i ti ri ti ti ti ii ri ri ri ri ri ri ri 11 n 11 ti ft CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE AGUA HEDIONDA LAGOON , CA APRIL 1994 0.5 ,- 1000 • CLOSURE PROBABILITY NORMAL FLOW ADJUSTED PLOW 10 15 20 TIME (DAYS) FIGURE 41 Closure probability for the month of April during a biannual dredge cycle II II II II II II I 1 I 1 fill I 1 II f 1 II 11 II 11 fill HISTOGRAM OF CLOSURE PROBABILITIES FOR A TYPICAL BI-ANNUAL DREDGE CYCLE C/3 > < Q &Ho KW PQ 120 100 80- 60 -- EXISTING CONDITIONS 40- 20-- 0.0 0.1 0.2 0.3 CLOSURE PROBABILITY 0.4 FIGURE 42: Histogram of closure probabilities for existing bathymetry during a typical biannual dredge cycle. t i r i t I f i t i t 1 I l I 1 i 1 i i I i i i I i i i i i i i i CLOSURE PROBABILITY DISTRIBUTION FUNCTION COoI-Jo En O ^ C/3 0 CJ EXISTING CONDITIONS 0 30 60 90 120 150 180 210 240 270 300 330 360 TIME ( DAYS ) FIGURE 43: Accumulated risk of inlet closure for existing bathymetry during the first year of a biannual dredging cycle. r i r i 11 11 f i t i ii ii i i ii ii 11 i i ii 11 ii 11 ii if CLOSURE PROBABILITY DISTRIBUTION FUNCTION DURING A TYPICAL BI-ANNUAL DREDGE CYCLE o_Jo Pt,o w OQI— I K Q o EXISTING CONDITIONS 0 100 200 300 400 500 600 700 TIME ( DAYS ) FIGURE 44: Accumulated risk of inlet closure for existing bathymetry throughout a biannual dredging cycle. 79 VI) ANALYSIS OF POTENTIAL REMEDIAL MEASURES TO REDUCE INLET CLOSURE RISKS Three distinctly different approaches for reducing inlet closure risks are studied. Each of these approaches seeks to reduce inlet closure risks by achieving increases in the net tidal prism. The remedial approaches studied are: A) manipulation of plant flow rates to reduce plant consumption of available tidal prism during periods of high risk; B) modification of choke point geometries to increase net tidal prism by reducing head losses, and C) dredging the entire lagoon system in a four-stage dredging plan to reduce both static and dynamic losses of tidal prism. Each of these remedial approaches is evaluated by repeating the tidal hydraulics and closure analyses for the tides, waves, and plant flow rates of the surrogate year from July 27, 1993 through July 27, 1994. The resulting closure risk statistics for each remedial method are compared against the results for the existing lagoon given in Section V in order to determine relative effectiveness. The relative effectiveness of each remedial method is weighed against other considerations such as operational compatibility with plant generating requirements, prior experience, cost and ease of implementation. These factors taken together are used to determine the best remedial approach, which is then restudied for the extreme event tides and mean tides to determine potential environmental benefits. A) Manipulation of Plant Flow Rates When equation 31 yields closure probabilities in excess of 10% for a particular time, t = t0, the computational routine enters a series of logical if-statements. These if- statements sequentially reduce the plant inflow rates according to a prescribed rate reduction schedule, and rerun the tidal hydraulics model for each flow rate on that schedule until a closure probability less than 10% is achieved. If no amount of flow rate reduction lowers closure probabilities below 10%, then a second series of if-statements will increase plant inflow rates according to the same cascade of flow rates. In either case, the point of entry into the cascade is that flow rate on the schedule closest to the actual plant flow rate which yielded the closure probability in excess of 10%. The cascade for rate changes is dictated by the various combinations of pumps which may be either brought on or off-line to vary the plant inflow rates between maximum and minimum capacity. The plant inflow rate cascade for rate schedule reductions or increases is shown in Table 2. 80 TABLE 2 PLANT INFLOW RATE CASCADE (ft3/sec) 1,250.2 1,196.6 1,143.2 1,089.7 1,036.3 982.8 929.3 706.5 483.7 252.0 20.26 (mgd) 808.0 773.4 738.9 704.3 669.8 635.2 600.6 456.6 312.6 162.9 13.10 The deviations from normal plant operations that would be necessary to reduce closure probabilities below 10% at any given time, t = tos are summarized over the calendar year in Figure 45. Inspection of Figure 45 reveals that 180 separate manipulations of the plant inflow rate schedule would be necessary to prevent the 361 days of Pc (O > 0.1 from occurring over a two-year dredge cycle. In the plots of the monthly closure probabilities, Pc (t0) = fc (t0) At, shown in Figures 39-41, the deviations from the normal flow curve that were necessary to reduce elevated closure probabilities below 10% are shown as green triangles labeled as "adjusted flow." In the black and white monthly plots of Appendix n, the normal plant flow rates are the dashed-dot-dot line; and deviation from normal flow necessary to reduce closure probabilities below 10% are the dashed spikes projecting below or above the normal flow rate curve. Correlation analysis of the flow rate deviations in Figures 39-41 and Appendix n revealed that every deviation from normal activity was required during the ebb tide intervals only. Furthermore those deviations requiring flow rate increases to drive the lagoon flushing ri r~i n r*i'r~i r~i n i i r ..... t ri i t i i i i i i i i SDG&E ENCINA POWER PLANT OUTFALL FLOW RATE JULY 27, 1993 - JULY 27, 1994 1000 r- 800 - 600 = 400 - 200 0 0 50 1:00 P.M. JULY 27, 1993 • ACTUAL PLANT FLOW RATE DEVIATIONS IN FLOW RATE A A AA AAAAAA AAA 100 150 TIME ( DAYS ) 200 250 300 350 1:00 P.M. JULY 27. 1994 FIGURE 45: Plant flow rate time history with deviations necessary to reduce closure probabilities below 10% r^ mm ri «FI rii ri rr ri r*i t i i I* SDG&E ENCINA POWER PLANT FLOW RATE HISTOGRAM JULY 27, 1993 - JULY 27, 1994 20 r- DEVIATIONS TO NORMAL OPERATION 15 10 8 I PLANT OUTFALL FLOW RATE (mgd) FIGURE 46: Histogram of altered flow rates necessary to reduce closure probabilities below 10% ii ii t I If II i I • I i 1 tl • • i I f I • i i 1 I I II I ! ill! C/Q 288 -- 264 240 216 O 192 + Cn 168 144 + 120 96 -- 72 -- 48 -- 24-- o CQ S D HISTOGRAM OF PLANT FLOW RATE ADJUSTMENTS FOR LESS THAN 10% CLOSURE PROBABILITY n *t f-4-0 -1000 -750 -500 -250 0 250 500 750 PLANT FLOW RATE ADJUSTMENTS ( mgd ) FIGURE 47: Histogram of incremental flow rate adjustments necessary to reduce closure probabilities below 10% 1000 r i f 1 II II n R 1 K 1 »;l f I f t I 1 r 1 PI II fill HISTOGRAM OF CLOSURE PROBABILITIES FOR A BI-ANNUAL DREDGE CYCLE WITH ADJUSTED PLANT FLOW RATES o P^Ho w K H PQ 0 0.00 0.01 0.02 0.03 0.04 0.05 0.06 0.07 0.08 0.09 0.10 CLOSURE PROBABILITY FIGURE 48: Histogram of closure probabilities resulting from flow rate deviations per Figure 21 r t t i r i t i t j 1 1 11 t • g i t i • t i i i i i i i i i § i o CUMULATIVE CLOSURE PROBABILITY BI-ANNUAL DREDGE CYCLE &-H 1— < g 0.025- OQ O g 0.020- K ^ 0.015- O H-J O ^ 0.010- ^ 0.005- o n nnn FOR ADJUSTED PLANT FLOW RATE .' i i i 100 200 300 400 500 600 700 TIME ( DAYS ) FIGURE 49: Accumulated risk of closure for existing bathymetry using plant flow rate adjustments throughout a biannual dredging cycle. 86 backwards (inflow during ebb tide) coincide with neap tides. Flow rate increases had to reach 739 mgd in order to be effective in dropping closure probabilities below 10%. A histogram in Figure 46 summarizes the numbers of operating days that plant inflow rates would have to be manipulated in a typical operating year. A total of 47 operating days, or 1,128 operational hours spread out over 180 separate manipulations of the plant flow rate would be required to keep closure probabilities under 10% on a daily basis. However, these 180 manipulations only involve four different settings (13.1, 162,9, 312.6 and 738.9 mgd) out of the eleven possible selections from the plant inflow rate cascade in Table 1. Unfortunately more than half of these manipulations involve deep cutbacks to either 13.1 or 162.9 mgd. The histogram in Figure 47 shows how flow rate manipulations were divided among decreases vs. increases in terms of number of operating hours. Most decreases in flow rates involved cuts of about 250 mgd below normal operating levels. The most common flow rate increases were of the same order. Although the required deviations in plant inflow rates are both numerous and sever, they do have a dramatic effect in reducing closure probabilities, Pc (tj. The histogram in Figure 48 shows that these flow rate manipulations allow no value of Pc(tJ to ever exceed 7.8% throughout a biannual dredge cycle, and that the great preponderance of closure probabilities remain under 3 %. This greatly reduced the accumulated risk. Referring to Figure 49, the closure probability distribution function using the adjusted plant flow rate schedule is plotted as a dotted line. With inflow rate manipulation, the accumulated risk of closure reaches only VA% after the first year of a biannual dredge cycle, and only 2Vz % at the end of the dredge cycle. At that rate of increase, it would take 40 years before inlet closure would become more probable than not. B) Modification of Choke Points The lagoon design has always suffered certain inefficiencies associated with frictional head losses in the three major choke points of its layout, namely: 1) the ocean inlet, 2) the railroad bridge, and 3) the 1-5 bridge. These head losses represent the flow energy which is lost to friction when the high speed flow through the choke points decelerates into the expanding portions of any of the lagoon basins. As a result of this loss of energy, not all of the velocity head in the choke point (kinetic energy) gets reconverted into pressure, i.e., tidal elevation (potential energy) within the lagoon basin. Thus, the frictional head losses of the 87 choke points equate directly with a loss of tidal range within the lagoon. The loss of tidal range in turn results in a loss of tidal prism which ultimately aggravates closure tendencies. Therefore, two questions arise: 1) Are choke point head losses big enough to compensate for the static and dynamic losses of tidal prism due to plant inflow rates and sediment infilling? 2) Can enough of these head losses actually be recovered through existing hydrodynamic technology to make a significant reduction in the risk of inlet closure? To answer the first question, the waterlines for spring, mean and neap tides were analyzed (see Figures 18, 19, 21, 22, 24, and 25). The decline in tidal range across each choke point was taken as a measure of the total head losses over a tidal cycle. Those head losses accumulate over all three choke points to produce a total tidal range loss at the east end of the east basin. The head losses factored over the various areas of each basin give an estimate of the prism losses. These prism losses due to the choke points are expressed as a percentage of the 1954 post-dredging configuration of the lagoon. The results are summarized in Table 3 below. TABLE 3 CHOKE POINT HEAD AND PRISM LOSSES OCEAN INLET HEAD LOSS RAILROAD BRIDGE HEAD LOSS 1-5 BRIDGE HEAD LOSS TOTAL TIDAL RANGE LOSS CHOKE POINT PRISM LOSSES PRISM LOSSES SINCE 1954 SPRING 0.97ft 0.46ft 0.62ft 2.05ft 16.9% 20.5% NEAP 0.51ft 0.22ft 0.31ft 1.04ft 25.6% 29.1% MEAN 0.83ft 0.41ft 0.56ft 1.80ft 27.1% 35.4% Inspection of Table 3 reveals that if the choke points could be made 100% efficient, then the equivalent of 27.1 % of the original mean tidal prism could be recovered. This recovery would be a recovery of potential tidal prism that was never realized due to design inefficiencies. But that recovery could be significant and would offset all but 8.3% of the 35.4% loss of original mean tidal prism due to sedimentation and increased plant flow rates. However, perfect choke point efficiency is not achievable. The decisive question then becomes #2 above; i.e., can enough head loss recovery be achieved through choke point reconfiguration to produce a significant reduction in inlet closure risk, even in the absence of any central and east basin dredging? The choke point optimization procedure is adapted to two physical constraints: 1) the minimum channel depth and width under the bridges must be preserved to prevent undermining existing bridge footings; and 2) disruption of existing eel grass beds must be avoided to ensure permittability. Both these physical constraints entail potentially large cost factors if violated. Constraint #1 defines the minimum channel cross-sectional area, A,,, for each choke point; and thus, the hydraulic control point and the velocity maximum for which an optimal pressure recovery solution must be found. The California coordinates for each of these hydraulic control points is as follows: TABLE 4 CALIFORNIA COORDINATES FOR HYDRAULIC CONTROL POINTS, a0 Ocean Inlet Railroad Bridge 1-5 Bridge X (ft easting) 1, 665,560 E 1,667,660 E 1,668,600 E Y (ft northing) 357,948 N 357,306 N 357,417 N Constraint #1 rules out the most straightforward approach to minimizing choke point head losses, namely, increasing the choke point channel cross-sectional area by either widening, deepening, or both. The conditions at the hydraulic control point determine how large a pressure recovery is necessary for a given choke point and over what distance it can be accomplished in an optimal way. The pressure recovery distance is the critical factor in flBm>89w» avoiding violation of constraint #2 above. ^w Ig There are two types of drag forces which perform dissipative work against the flow and thereby determine the loss of flow energy that manifests itself as head losses. These are: 1)^^w I* turbulent wall friction, and 2) form drag. Form drag results when the flow through a choke ^ point decelerates too rapidly into the next adjacent basin causing it to separate from the banks tm and form eddies. These eddies waste flow energy by doing work moving sediment into m unwanted shoals and bars and by dissipating flow energy into heat. Turbulent wall friction along w the channel side slopes and bottom produces similar wastes of flow energy in creating unwanted p* scour holes. The key to minimizing head losses through a choke point is in causing the flow ** to decelerate from a velocity maximum at the hydraulic control point, such that the turbulent *. flow does not separate and the wall friction is minimal. In this way, the maximum amount of ** kinetic flow energy is converted back into potential energy as pressure, i.e., optimal pressure jp, recovery. The rate of flow deceleration or pressure recovery with distance is the decisive ** process in achieving this optimal condition. This rate shall be referred to as the pressure *» recovery distribution. *"" B.S. Stratford (1958a and b) developed an analytic solution for a pressure recovery *•* distribution in a turbulent flow which both avoids flow separation while maintaining zero wall (w friction over the entire pressure recovery length. This solution is now widely regarded as a ** universal law for the minimum energy dissipation in a turbulent flow. It has been used to l*optimize turbine blades, lifting bodies, aircraft wings, ground effects tunnels and wings for race ** cars, flow nozzles and diffusers (see Liebeck, 1976; Smith, 1974; Wortmann, 1974; and to McCormick, 1979). The present application of the Stratford pressure recovery to a tidal inlet ** has never been attempted before, but is very similar to the prior applications to flow nozzles.»»The primary distinction is that tidal flow through a choke point is bi-directional, whereas prior nozzle applications have been made for unidirectional flows.t» Stratford recovery is sometimes referred to as "imminent separation recovery" because it works by keeping the flow on the verge of separation throughout the entire deceleration without actually separating. In this condition, the eddy formation is prevented while velocity n*shear at the wall is close to zero; and, consequently, the turbulent wall friction nearly vanishes. This is how most of the energy dissipation is avoided and how most of the kinetic energy of the 90 *> velocity head is reconverted into pressure. A typical Stratford pressure recovery distribution isPI shown in Figure 50. To avoid dependence on the exact pressure and velocity of a particular flow, the pressure recovery is expressed in terms of pressure coefficient, C P , defined as I* - P — PP- C = _— -2. (32) ** Here p is the density of the flow; u0 is the velocity of the flow at a point a = o0 where the ** pressure recovery is initiated; P0 is the static pressure at a = a0; and P is the local static '^m pressure at any point along the pressure recovery length o > a0. The canonical coordinate, a0, m corresponds to the hydraulic control points in Table 4, For reasons which date back to the initial application to airfoils, Liebeck and Ormsbee (1970), the pressure coefficient is plotted on ** an inverse scale with C P = 0 at the top of the axis scale, and positive values below. When MM ____ C p = 0, the flow is at its maximum velocity, u = u0, and is about to convert the velocity head, m Vipu o, back into pressure, whence C P > 0. If the flow reconverts all of its velocity head into »pressure (total pressure recovery), then the flow would be brought to rest, u0 = 0. This is a ** condition known as stagnation for which the pressure coefficient has a maximum value of C P•» = 1.0. Total pressure recovery will not occur in the present application because the flow can not be brought to rest within a basin and still be able to pass water between basins as the tidal elevation varies. Therefore, the pressure recovery distribution in Figure 50 never reaches CP = 1.Mat • _ The most distinctive feature of the Stratford recovery is the initial sharp increase in C P•*» (deceleration) immediately after the pressure recovery begins at a = a0. This places the flow on the verge of separation; but the initial flow deceleration is not made so large as to actually te induce separation. Further flow deceleration and C P increases are made progressively more and more gradual in the downstream direction. This produces a concave pressure recovery 40 distribution known as a "pressure recovery bucket." Stratford (1958) proved that the concave pressure recovery distribution had a stabilizing influence on a decelerating flow that was close i > ft t i f i tt i t i f i • i 11 11• i tt ti i i CANONICAL FORM FOR STRATFORD PRESSURE RECOVERY DISTRIBUTION o Cn= 0 0u MAXIMUM VELOCITY, u = UQ oo DQ CO OH + 1.0 STAGNATION, u = 0 O d. Ou u Q >- ffi PRESSURE RECOVERY cr = cr 0 CANONICAL CHANNEL LENGTH, (7 FIGURE 50:Schematic of the Stratford Pressure Recovery Distribution. 92 to separation. It was found to be the hydrodynamic equivalent of "having your cake and eating it, too"; i.e., eliminating wall friction by imminent separation while preventing separation with a stabilizing concave pressure recovery distribution. In addition to providing the maximum possible pressure recovery, the Stratford distribution also provides that recovery over the shortest possible distance, a,. The Stratford pressure recovery relations are implemented in canonical coordinates, a, which in the present application are curvilinear along the channel axis. The canonical coordinates for each choke point are solved from digitized files of bank coordinates. These bank coordinates are expressed with respect to California coordinates. For every X (easting), there is a corresponding north bank, Y = Yn, and south bank, Y = Ys. The canonical coordinate, a, is given by a finite difference solution to da2 = dX,dYn,dY. Sxx SXN Sxs SNX SNN SNS 8sx SSN Sss dX (33) where gy is the metric tensor which represents the coefficients that arise in computing for any depth contour dX2 + dY2 + dZ2. The finite difference step interval used in solutions to Equation (33) was 20 ft, with 51 such computational steps used to resolve the ocean inlet and railroad bridge choke points, and 39 for the 1-5 bridge. With the canonical coordinate at the hydraulic control points, tr0, and the velocities at those points, u0(t), the choke point Reynolds number may be defined as «w r R«(t) =(34) where v is the kinematic viscosity. With the terms u0(t), a0, R^t), the optimal Stratford pressure recovery distribution can be specified for each choke point and each tide. The Stratford distribution has an inner and outer solution domain which overlaps at C P ~ 4/7. During the initial, rapidly decelerating portion of the distribution in the neighborhood of the hydraulic control point, C P ^ 4/7, the pressure recovery distribution is given by (see Stratford, 1958), 93 Cp(<r/<r0) =0.49 1 for C ^l'-ll•• J (35) 4/7. In the later stages of the pressure recovery as the flow is decelerating to stagnation, C P > 4/7, the distribution is given by Cp (cr/ac) = 1 -(36) for 4/7 < CP < 1. Here "a" is an integration constant chosen to smoothly match Equation (36) with Equation (35) in the overlap domain of C~ P ~* 4/7; and "b" is selected to define how close recovery is allowed to approach stagnation. Equations (35) and (36) are then used to specify the optimal velocity distribution U (a, t) as a function of the canonical length along the channel axis, a. From the definition of pressure coefficient from Equation (32), we write: U(M) = uc(t)(37) The minimum distance, a,, over which the pressure recovery can take place according to Equations (35) and (36) is given by a. = 94 1 f .06b io-( dCpli L. i i ^ 5 Re(t) Cp 1 1C •I (a - 2)/4 (38) The optimal velocity distribution for all three choke points during flood flows due to spring, neap, and mean tides are calculated from Equations (35) through (38) and are plotted in Appendix ffl. The corresponding ebb flow optimal velocity distributions are plotted in Appendix IV. No optimal ebb flow distribution was possible for the ocean inlet due to the jetties, which were included under constraint #1 above. Figure 51 shows the optimal velocity distribution for the ocean inlet during a flooding spring tide. The figure also includes the existing velocity distribution plotted as a dashed line. Both are plotted against distance along the channel centerline beginning from the control point as measured into the west basin with the flood flow. Note how the existing channel velocities initially decelerate too slowly. The flow in this region immediately downstream of the control point has an excess velocity head; and it is here along the first 60 meters where the inlet channel presently scours. Beyond 60 meters channel distance, the existing flow decelerates too rapidly and thus has insufficient flow energy. It is in this region between 60 and 240 meters downstream from the control point where the large inlet flood bar forms, and where most of the recharge volume is filled. Similar behavior is found in Figure 52 for flooding spring tides under the railroad bridge. A long reach of sub-optimal flow extends from the control point into the central basin and has a corresponding sand bar deposit that has never been dredged (see Figure 5). Figure 53 reveals that velocities under the 1-5 Bridge initially decelerate too slowly into the east basin during flood tide. In this region beyond the rock scour blanket under the bridge, deep scour holes have been found. Beyond the first 100 meters into the east basin, the velocities presently decelerate much too rapidly and the large east basin sand bar has shoaled and subsequently vegetated with eel grass as shown in Figure 6. Similar deviations from optimal are found in the existing velocity distributions during ebb tide, as shown in Figures 54 and 55. The ebb flow velocity distributions are found to the left il c f i it ti ti ft ft € i t 1 I i ii it ii if i i ii i FLOOD FLOW VELOCITY DISTRIBUTION FOR SPRING TIDE MAXIMUM OCEAN INLET, AGUA HEDIONDA LAGOON, CA 200 o CD ao E->—i O O 180- 160- 140- 120 100 80'I 5o(V. 60 + 40 + 20 0 CJ KQ OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 0 30 60 90 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) FIGURE 51:Optimal vs. existing flood flow velocity distributions at the ocean inlet for a spring tide maximum. r l I t t t I f i f i f i f i f i •i i i t t t i t t i i i i i FLOOD FLOW VELOCITY DISTRIBUTION FOR SPRING TIDE MAXIMUM RAILROAD BRIDGE, AGUA HEDIONDA LAGOON, CA 200 o CD W ao E-i—i O O 180 160- 140- 120- 100- 80- 60- 40 20 + 0 OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS O D-, •JO KE- & Ou CJ 0 30 60 90 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) FIGURE 52:Optimal vs. existing flood flow velocity distributions at the railroad bridge for a spring tide maximum. r i t i t r t i i r t i i i t f i f i i i i i i i i i i i i i i i FLOOD FLOW VELOCITY DISTRIBUTION FOR SPRING TIDE MAXIMUM o CDin eo E--i—i O O 200 180 + 160 140 120 100 80 60 40 20 + 0 1-5 BRIDGE, AGUA HEDIONDA LAGOON, CA O O, JO KH2 Ou u KQ OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 0 20 40 60 80 100 120 140 160 180 200 220 CHANNEL CENTERLINE DISTANCE ( m ) FIGURE 53:Optimal vs. existing flood flow velocity distributions at the 1-5 bridge for a spring tide maximum. ii r i t i ft c I t i t I t i r i r t i i i i i i t i i i t i i t i o eo oo EBB FLOW VELOCITY DISTRIBUTION FOR SPRING TIDE MAXIMUM 200 180 0 RAILROAD BRIDGE, AGUA HEDIONDA LAGOON, CA OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 60 90 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) FIGURE 54:Optimal vs. existing ebb flow velocity distributions at the railroad bridge for a spring tide maximum. i i f i t i i i rtfifififirifiiiiiiitiiiiiii o CD E O Oo EBB 200 180 160 140 120 100 80 60 40 20 0 FLOW VELOCITY DISTRIBUTION FOR SPRING TIDE MAXIMUM 1-5 BRIDGE, AGUA HEDIONDA LAGOON, CA E- I— I O CL OCJ o KQ OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 0 20 40 60 80 100 120 140 160 180 200 220 CHANNEL CENTERLINE DISTANCE ( m ) FIGURE 55:Optimal vs. existing ebb flow velocity distributions at the 1-5 bridge for a spring tide maximum. 100 of the control point, measured with canonical length along the channel centerline axis. Inspection of the remaining velocity distributions in Appendices HI and IV indicates that none of the choke points are even close to optimal in their pressure recovery velocity distributions. Consequently, the very large head losses and reductions in tidal prism evident in Table 3 have resulted. Note that none of the optimal distributions in Figures 51-55 is allowed to drop to stagnation. This is due to the ambient pass-through flow in the basins during the tide cycle. Figures 56-58 give the composite optimal velocity distributions for both flood and ebb flow for all tides and choke points. Note that the pressure recovery distances, a,, are not symmetric for flood vs. ebb. This was in part due to the selection of the parameter b in Equation (36) in order to minimize encroachment on the eel grass beds. These asymmetric velocity distributions will dictate the shape of the choke point reconfigurations, which will in turn become asymmetric structures. Included in Figures 56-58 is a horizontal line indicating the threshold velocity of the mean size fraction of sediment found in the lagoon according to Table 1. The threshold sediment transport velocity will truncate the optimal velocity distributions because sediment deposition will occur in the sub-threshold portions of the recovery length. That sediment deposition will ultimately shoal the bottom and elevate the velocity to near threshold levels, no matter what intervening structure is present. Therefore, choke point reconfigurations will not entail structures longer than the recovery length dictated by the intersection of the optimal velocity distribution and the threshold velocity. It is noted that much of the neap tide optimal velocity distributions remain sub-threshold. Thus, the inverse problem for the intervening choke point structures was weighted more heavily towards the mean and spring tide velocity distributions. The flood tide distributions appear on the righthand side of the control points in Figures 56-58 while the ebb tide distributions are on the left hand side. Consequently, there will be right hand and left hand solutions for the shape of the structures posed for each choke point. i i f i if i f 1 I 'f » i f i i f i i i i i i i i t i i i i i t I o CD 00 eo o CANONICAL STRATFORD VELOCITY DISTRIBUTION OCEAN INLET, AGUA HEDIONDA LAGOON, CA 200 180 160 140 120- 100- 80- 60 40- 20-= 0 EXTREME SPRING TIDE MEAN TIDE EXTREME NEAP TIDE SEDIMENT THRESHOLD VELOCITY 0 30 60 90 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) FIGURE 56:Family of optimal velocity distributions at the ocean inlet for all tidal ranges. r i t i f i 1 i r i r r i t 1 f i i i i i i t i i i i i i i I t cj CD so E-"i—i O Oh-J 200 CANONICAL STRATFORD VELOCITY DISTRIBUTION RAILROAD BRIDGE, AGUA HEDIONDA LAGOON, CA EXTREME SPRING TIDE MEAN TIDE EXTREME NEAP TIDE SEDIMENT THRESHOLD VELOCITY 0 30 60 90 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) FIGURE 57:Family of optimal velocity distributions at the railroad bridge for all tidal ranges. f 1 11 f I 111 i r i i i i i i i f i i f i i i i i i i i t i i i i i o CD oo CANONICAL STRATFORD VELOCITY DISTRIBUTION 1-5 BRIDGE, AGUA HEDIONDA LAGOON, CA EXTREME SPRING TIDE MEAN TIDE EXTREME NEAP TIDE SEDIMENT THRESHOLD VELOCITY 0 20 40 80 100 120 140 160 180 200 220 CHANNEL CENTERLINE DISTANCE ( m ) FIGURE 58:Family of optimal velocity distributions at the 1-5 bridge for all tidal ranges. 104 We now seek solutions to the inverse problem for the physical shapes of the choke points which will yield the optimal velocity distributions summarized in Figures 56-58. From continuity principles, it is apparent that the choke point cross-sectional areas, A (a, t), must vary along the pressure recovery lengths according to: (39) where A,, is the minimum cross-sectional area at the hydraulic control points, a = a0. However, there are an infinite number of possible combinations of depth variations and channel width variations that could yield the cross-sections given by Equation (39). The number of such choices is narrowed considerably by applying constraint #2 outlined above. The number of possible width and depth choices is further reduced by imposing a third constraint: 3) that deviations to the present lagoon geometry are minimized. This constraint is imposed due to considerations of constructability and cost minimization. From these constraints, two classes of solutions have emerged: A) a constant depth solution referred to as the "Stratford Fence," and B) a variable depth solution, referred to as the "Stratford Bottom." Both of these solutions require deviations to the present channel widths through the choke points. These width deviations are accommodated by vertical flow fence geometries. The constant depth solutions (the Stratford Fence) are developed by selecting a depth over the entire pressure recovery region that is based upon the present structurally hardened depth at the hydraulic control points, Z0. These depths insure the structural integrity of existing bridge footings. A depth of Z0 = —10.0 ft NGVD was selected for the pressure recovery region of the ocean inlet based upon the existence of a concrete pad at that depth under the Highway 101 bridge. Basement depths of Z0 = -9.0 FT NGVD and Z0 = -9.5 ft NGVD were selected for the railroad and 1-5 bridges, respectively, due to existing rock work under those bridges. By fixing the depth for any given choke point, the optimal velocity distribution is achieved with channel width variations using a flow fence. The flow fence geometry is adapted to the existing lagoon geometry by taking the transpose of the metric tensor, g^, , used in the conversion of the bank coordinates (X, Yn, Ys) to canonical coordinates, a. Since the channel 105 width set by the flow fence must be A (a, t)/Z0, the flow fence coordinates for the flat bottom solutions are given by (40) where g T is the transpose of the metric tensor. Solutions to Equation (40) involve a right hand solution for the flood flow optimal velocity distributions and left hand solutions for the ebb flow distributions. Appendix V gives the California coordinates for the flow fences of the flat bottom solutions, (X, Yn*, Y,* ), or Stratford Fences. These are plotted in Figures 59-61 relative to the MHHW contour (red) and the MLLW contour (blue). Each choke point is modified from its present configuration by a constant depth hard bottom laid between a pair of north and south bank flow fences. The hard bottom consists of a layer of AnnorFlex or rock. Rock bedding should be nominally 1.0 ft quarry stone placed in a layer 3.0 ft thick. A total of 792,000 ft2 of rock or AnnorFlex bedding is required for this concept. The flow fences are nominally sheet pile set to an elevation of +5.0 ft NGVD. Inspection of Figures 59-61 reveals a limited degree of cutting into the existing bank. The ocean inlet, Figure 59, requires the greatest modifications to the existing bank, but primarily the south bank. The north bank of the ocean inlet channel is essentially unchanged from its present configuration. The regions between the flow fences and the existing banks is free-flooding. Sheet pile joints should be sealed to prevent pressure losses due to leakage and ventilation. The second class of solutions involves an attempt to achieve the optimal Stratford velocity distribution within the existing banks, at least without cutting into the existing banks. For these solutions (the Stratford Bottom), a depth variation along the channel length, Z (X, Y, t), is calculated from the optimal channel cross-sections, A(<r, t), according to Z(X,Y,t) = i (41) g"T o AREA = 417,596 ft SHEET PILE FLOW FENCE i i i .1 i i i i i i i 0 100 200 300 400 500 SCALE IN FEET 1,666.241 E 357,035 N SHEET PILE FLOW FENCE ROCK/ ARMOR FLEX BEDDING 10.0 ft NGVD SCOTT A JPWff CUM9UUM7 OCEAN INLET RECONFIGURATION / FB 9CBMBR: SCOTT i. JIMCBB PhD ud JOSEPH WASH. FIGURE 59: Row fence and bottom detail for the flat bottom "Stratford Fence" at the ocean inlet SHEET PILE FLOW FENCE ROCK/ ARMOR FLEX BEDDING -9.0 ft NGVD I I 1 I I ' I 0 100 200 300 400 500 SCALE IN FEET 1,667,080 E 357,265 N 222,124 ft 1 I SHEET PILE FLOW FENCE RAILROAD BRIDGE RECONFIGURATION / FB SOBHWS scow i jams M> ud mm nan. FIGURE 60:Flow fence and bottom detail for the flat bottom "Stratford Fence" at the railroad bridge. SHEET PILE FLOW FENCE 1,669,000 E 357,644 N SHEET PILE FLOW FENCE i i i I _ I _ I 0 100 200 300 400 500 SCALE IN FEET ROCK / ARMOR FLEX BEDDING -9.5 ft NGVD AREA = 153,024 ft2 OR SOOtt A JRQBK? COHKVSK 1-5 BRIDGE RECONFIGURATION / FB scon i JBOEDB no at joamum FIGURE 61:Flow fence and bottom detail for the flat bottom "Stratford Fence" at the 1-5 bridge. I I I I I I I I I I I I I I I I I I 109 where g~T is the transpose of the metric tensor. However, some of the solution points calculatedby Equation (41) rise above the MLLW level, and would tints cut off tile flow before reaching MLLW. These solutions are truncated at a minimum depth 2^* = -5.0 ft NGVD to allow at least a foot of water depth for flow at the MLLW elevations. In this truncated case, the Stratford velocity distribution is continued with channel width variation using a flow fence. These width variations yield auxiliary flow fence coordinates, (X, Ya', Y'J, based upon the Z^ limit, or l g-*, (42) Solutions for the bottom profile along the length of each choice point channel are plotted in Figures 62-64. AH of the solution points of the ocean inlet channel in Figure 62 are below -5.0 ft NGVD, and hence no auxiliary flow fence is required there. However, the variable depth solutions for die railroad bridge, Figure 63, and the 1-5 bridge, Figure 64, must be truncated to avoid choking the minus tide flows. The California coordinates for the auxiliary flow fences in the truncated solution domains are tabulated in Appendix VI, and are plotted for the multiple basin system in Figure 65. In all cases, the minimum hardened depth of the choke ,vpoint has been preserved at the hydraulic control point. A total of 903,506 ft2 of ArmorFlex or rock bedding is required for all three choke points, 111,506 ft2 more than the constant depth solution. If rock bedding is selected, it is recommended that 1.0 ft quarry stone laid in a 3.0 ft deep layer be used. The variable depth plan, however, involves fewer flow fences (none for the ocean inlet) which are shorter. Both the flat bottom and variable depth concepts wfll alter sand deposition within the lagoon. Because nearly all of the pressure recovery length that has been hardened will have velocities in excess of the threshold velocity of the sediments, no sediment deposit will occur>r -in these regions. Consequently, the primary deposition area and recharge zone in the west lagoon will be displaced inward, and will not begin until me toe of the hardened ocean inlet section. Similar displacements are anticipated for the central and east basin. 1 I 1 t I t , f 1 f 1 I 1 f 1 f 1 I i I 1 I 1 I 1 f I f I I I I I I I Q O -4-> ^4-H O o PQ E- Q BOTTOM VARIATIONS FOR STRATFORD PRESSURE RECOVERY OCEAN INLET, AGUA HEDIONDA LAGOON, CA 20 18 16 14 12 10 8 6 4 2 0 oCO IDin CD CD X o OH STATFORD RECOVERY BOTTOM PROFILE 0 100 200 300 400 500 600 700 800 900 1000 CHANNEL CENTERLINE DISTANCE ( ft ) , SECTION A-A FIGURE 62:Variable depth solution to the "Stratford Bottom" for the ocean inlet. f i i i i i r i f i i r i r i i i i i r i i i i i i i t i Q O -^ «+-! O o OQ Q BOTTOM VARIATIONS FOR STRATFORD PRESSURE RECOVERY RAILROAD BRIDGE, AGUA HEDIONDA LAGOON, CA 20 18- 16- 14- 12- 10- 8- 6 4 2 + 0 STATFORD RECOVERY BOTTOM PROFILE STRATFORD RECOVERY BOTTOM / FENCE o COCDr-CDCD X H 55I—tOCX 55OO u KQ 0 100 200 300 400 500 600 700 800 900 1000 CHANNEL CENTERLINE DISTANCE ( ft ) , SECTION B-B FIGURE 63:Variable depth solution to the "Stratford Bottom" for the railroad bridge. i r i t i i i i i r i i i i r i i r i t i i i i i t i t i f i i i i § BOTTOM VARIATIONS FOR STRATFORD PRESSURE RECOVERY 1-5 BRIDGE, AGUA HEDIONDA LAGOON, CA pn<j\j 18- g 16- C555 14- o 12' 3 10~W p PQ 0- r™*^ ^"i^ b - P-H Q 4- 2- n . \^ / \/ \ / ^ «E-< ^*^-«y •~H D 05 Q W STATFORD RECOVERY BOTTOM PROFILE ----- STRATFORD RECOVERY BOTTOM / FENCE /"" \/ \ * \ ,.-'"' \ / I ; \ 8 1 § 1 FLOW FENCE CD 1 / X \ ^^^^ 0 100 200 300 400 500 600 700 CHANNEL CENTERLINE DISTANCE ( ft ) , SECTION C-C FIGURE 64:Variable depth solution to the "Stratford Bottom" for the 1-5 bridge. VARIABLE DEPTH ROCK / ARMOR FLEX AREA = 344,313 ft2 w< VARIABLE DEPTH ROCK / ARMOR FLEX AREA = 296,028 ft VARIABLE DEPTH \\ :OCK / ARMOR FLE)C\ AREA = 201,355 ft/« 1,667,900 E 357,838 N SHEET PILE FLOW FENCE CHOKE POINT RECONFIGURATION / VB JUUIISU: 300TT A. JBOCBB PhD tad J08VB USYL FIGURE 65:Multiple basin choke point modifications for the variable depth "Stratford Bottom" solution. 114 To evaluate the tidal prism recovery and improvement to inlet closure risk, the numerical tidal hydraulics model was reprogrammed and initialized to accept the flat bottom and variable bottom choke point modifications. Both concepts were tested in continuous one-year simulations using the 1993 bathymetry and the corresponding flow rate, wave and tide data for the surrogate year of July 27, 1993 to July 27, 1994. The achieved hydraulic improvements are summarized in Table 5, in which the flat bottom concept that relies only on flow fences is referred to as "Stratford Fence" and the variable bottom concept is referred to as "Stratford Bottom." (See Table 5 on following page.) TABLE 5: SUMMARY OF CHOKE POINT RECONFIGURATION *NORMAL PLANT INFLOW RATE HISTORY July 27, 1993 through July 27, 1994 115 1 I. TIDAL RANGE LOSSES DUE TO CHOKE POINTS Spring Neap Mean Existing Conditions 2.05ft 1.04ft 1.80ft Stratford Fence 0.812 ft 0.450 ft 0.778 ft Stratford Bottom 1.092 ft 0.583 ft 1.01ft H. TIDAL RANGE RECOVERY Spring Neap Mean Stratford Fence 1.238 ft 0.590 ft 1.022 ft Stratford Bottom 0.958 ft 0.457 ft 0.790 ft HI. TIDAL PRISM LOSSES DUE TO CHOKE POINTS (% OF DESIGN TIDAL PRISM) Spring Neap Mean Existing Conditions 16.9% 25.6% 27.1% Stratford Fence 6.69% 11.08% 11.7% Stratford Bottom 9.00% 14.35% 15.20% IV. TIDAL PRISM RECOVERY (% OF DESIGN PRISM) Spring Neap Mean Stratford Fence 10.21% 14.52% 15.40% Stratford Bottom 7.9% 11.25% 11.90% I* Jug im tm m ID* f* 1 • T* In in i • I I I I I I I I fi I I I I I I I I I 116 Inspection of Table 5 reveals some significant hydraulic improvements achieved by either concept, although the flat bottom "Stratford Fence" is slightly better and involves less rock or ArmorFlex work. Mean tidal range losses were improved from 1.80 ft to only 0.778 ft for the "Stratford Fence," and 1.01 ft for the variable depth "Stratford Bottom." As a result, 15.4% of the mean tidal prism was recovered by the "Stratford Fence" concept, while 11.9% was recovered by the "Stratford Bottom" concept. This prism recovery represents a large fraction of the 27.1 % of the mean potential tidal prism lost to choke point inefficiencies; and makes a substantial compensation for the 35.4% loss in mean design prism that has resulted since 1954 due to sedimentation and plant expansion. The resulting instantaneous closure probabilities for the summer and winter months that had previously been the worst-case scenarios are shown in Figures 66 and 67. Inspection of these figures show that each and every episode of enhanced closure risk was diminished by either concept, with the constant depth "Stratford Fence" slightly better than the variable depth "Stratford Bottom." The accumulated risk from these effects factored over a typical 2 year dredge cycle are shown in Figure 68 and summarized in Table 6. TABLE 6: PROBABILITY OF INLET CLOSURE Cumulative Closure Probability Closure Recurrence Interval Existing Conditions 11%/yr 4.5 yr Stratford Fence 5%/yr 10.0 yr Stratford Bottom 6%/yr 8.3 yr It is seen that the accumulated risk of closure has been diminished from 11 % per year to 5 % per year for the "Stratford Fence" and 6% per year for the variable depth "Stratford Bottom". Consequently, the inlet closure interval is more than doubled over present conditions by the Stratford Fence, and increased 84% over present conditions by the "Stratford Bottom". It is remembered that these improvements are calculated for the existing partially shoaled central and east basin bathymetries. Even greater improvement can be expected if any degree of reconstruction dredging was performed in the central and east basin concurrent with these choke point modifications. I PQ <PQ O H (73O-J O CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE AGUA HEDIONDA LAGOON, CA AUGUST, 1993 0.5 0.4-- PROBABILITY FOR NORMAL OPERATION - NORMAL FLOW RATE PROBABILITY STRATFORD RECOVERY BOTTOM - PROBABILITY STRATFORD. RECOVERY FENCE , 0.3-- 0.2 -- I, 0.1 — 0.0 14 21 TIME ( DAYS ) 28 -- 800.0 -- 600.0 --400.0 1000.0 -- 200.0 006 DH FIGURE 66:Instantaneous closure probabilities for the month of August during a biannual dredge cycle. CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE AGUA HEDIONDA LAGOON, CA FEBRUARY, 1994 DQ DH w o O 14 TIME ( DAYS ) -- 800.0 -- 600.0 1000.0 i --400.0 -- 200.0 FIGURE 67:Instantaneous closure probabilities for the month of February during a biannual dredge cycle. O55 Oh— HE- £>m COi— iQ PQ O «ou CLOSURE PROBABILITY DISTRIBUTION FUNCTION FOR CHOKE POINT MODIFICATIONS: BI-ANNUAL DREDGE CYCLE 0.20-r- 0.18- 0.16- 0.14- 0.12- 0.10- 0.08- 0.06- 0.04- 0.02- 0.00 STRATFORD PRESSURE RECOVERY FENCE STATFORD PRESSURE RECOVERY BOTTOM PROFILE 100 200 300 400 500 TIME ( DAYS ) 600 700 FIGURE 68: Accumulated risk of closure due to choke point modifications applied to existing bathymetry throughout a biannual dredge cycle. I I 1 I 1 t 1 t 1 t t t I t I I I t I 120 C) Reconstruction Dredging The reconstruction dredging remedial option seeks to reduce closure risks by enlarging the tidal prism and by removing existing choke point head losses associated with flow restricting bar formations like those shown in Figures 5, 6 and 10. When fully implemented, the dredge plan as laid out in Figure 69 will both recover a large portion of the original tidal prism lost to sedimentation, as well as recover an additional fraction of the potential tidal prism by means of more hydraulically efficient proportioning of intertidal acreage. The dredge plan will be implemented in four stages indicated schematically in Figure 70. The central features of each stage of the dredge plan are as follows (see Figures 69 and 70): STAGE 1: Removal of the middle basin sand bar to achieve a uniform depth of -8.0 ft NGVD in the middle basin. STAGE 2: Removal of the east basin sand bar by the 1-5 bridge to level the western section of the east basin at -8.0 ft NGVD. STAGE 3: Dredge a borrow pit to -19.0 ft NGVD in the central region of the east basin and fill it with 130,430 cubic yards of fine-grained sediments dredged from the mudflats and banks of the eastern end of the east basin. Dredge features in the eastern end of the east basin include general leveling to -9.0 ft NGVD; a triangle-shaped sediment trap lowered to -17.0 ft NGVD; and a channel to facilitate drainage of the Agua Hedionda Creek and neighboring mudflats with a bed slope of 9.4 to 1. STAGE 4: Removal of the inlet bar from the west basin to lower the recharge zone to —10.0 ft NGVD and progressively deepen the west basin to -20.0 ft NGVD toward the plant infall. Take 160,720 cubic yards of sand from west basin dredging activities and use it to cap the borrow pit in the east basin. FULLY IMPLEMENTED 4 STAGE DREDGE PLAN AGUA HEDIONDA LAGOON, CA MIDDLE BASIN: 38,000 cy EAST BASIN 253.150 cy CONTOURS IN FEET NGVD + 4 to -5; -9, -10, -13, -16 I 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET BORROW PIT CAPACITY: 291.150 cy EAST BASIN FINES: 130,430 cy SAND CAP 160.720 cy CHANNEL SLOPE = 9.4:1 SLOPE = 10:1 SEDIMENT TRAP DR SCOTT A. JENKINS CONSULTING TOTAL DREDGE VOLUME = 743,020 cy SCOTT A. JENKINS PhD & JOSEPH WASYL FIGURE 69: Layout of a four-stage reconstruction dredging plan. SEDIMENT PROBLEM REMEDIATION 4 - STAGE DREDGE PLAN 0 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA 648 mgd PLANT INFLOW SCOTT A. JENKINS PhD & JOSEPH WASYL FIGURE 70: Schematic of the sequence of stages hi the reconstruction dredging plan. I I Altogether, the four-stage dredge plan will remove 582,300 cubic yards of sediments • from the central and east basins, placing much of the sand-sized fractions on neighboring beaches while burying 130,430 cubic yards of fine-grained sediments in the east basin borrow I pit. A total of 253,150 cubic yards of sediments will be dredged from the east basin sand bars and mudflats, 38,000 cubic yards from the central basin sand bar system, and 291,150 cubic • yards of sand will be dredged from the east basin in creating the borrow pit. In addition, the west basin cleanup will dredge approximately 160,720 cubic yards used to cap the borrow pit, J bringing to total dredging volume for the fully implemented dredge plan to 743,020 cubic yards. Because the dredging production schedule will likely spread the implementation of the P various stages over several years due to permitting constraints, the tidal hydraulics of each stage was studied for end-to-end runs of the complete surrogate calendar year, July 27, 1993 through £ July 27, 1994. The resulting incremental probability density function, fvAV, for tidal prism ^ increments of AV = 1,000,000 ft3 are shown in Figures 71-74 for each stage of the dredge plan. Q Comparison of these figures with the tidal prism probability density function of the existing f lagoon shown in Figure 27 reveals how the net tidal prism progressively enlarges for each dredge stage under actual operating conditions. Stage 1 of the dredge plan is found in Figure — 71 to provide immediate and significant increases in net tidal prism, raising the mean tidal prism • by 6 million cubic ft and the maximum spring tidal prism by 12 million cubic feet. This I increase was achieved by only dredging 38,000 cubic yards, indicating how significantly the sand bars in the central basin have restricted flow into the east basin where the majority of the lagoon » system tidal prism resides. Stage 2 makes a much less dramatic improvement in net tidal prism under the operational conditions of the surrogate year (see Figure 72), improving the mean prism IB over Stage 1 by an additional 1 million cubic ft, and the maximum spring prism by an additional 3 million cubic feet. By Stage 3 (Figure 73), the mean prism is increased by 9 million cubic • ft over existing conditions while the maximum spring prism is increased by 19 million cubic ft over existing conditions. Upon completion of Stage 4, the mean prism has reached 40 million cubic ft under operational conditions (see Figure 74) while the maximum spring prism has been restored to the original design value of 80,000,000 cubic ft. The probability of the very small neap prisms is also greatly diminished by Stage 4 relative to the existing situation for the undredged bathymetry (see Figure 27). I 1 1 I _JI—IPQ PQ O OH TIDAL PRISM PROBABILITY DENSITY FUNCTION DUE TO STAGE 1 DREDGING: 27 JULY 1993 - 27 JULY 1994 6 5- 4-- o 1 " 0 0 m —f- 10 20 30 40 50 60 70 TIDAL PRISM ( MILLION CUBIC FEET ) 80 FIGURE 71: Incremental probability density function of the tidal prism for the stage 1 dredge plan during the period between July 27, 1993 and July 27, 1994. _qI—I PQ PQ O PL, TIDAL PRISM PROBABILITY DENSITY FUNCTION DUE TO STAGE 2 DREDGING: 27 JULY 1993 - 27 JULY 1994 6 5- 4-- 3- 1 " 0 0 10 20 30 4.0 50 60 TIDAL PRISM ( MILLION CUBIC FEET ) 70 80 FIGURE 72: Incremental probability density function of the tidal prism for the stage 2 dredge plan during the period between July 27, 1993 and July 27, 1994. oKPu TIDAL PRISM PROBABILITY DENSITY FUNCTION DUE TO STAGE 3 DREDGING: 27 JULY 1993 - 27 JULY 1994 6 5- 4-- 1 -- 0 0 IfflkJL 10 20 30 40 50 60 70 TIDAL PRISM ( MILLION CUBIC FEET ) 80 FIGURE 73: Incremental probability density function of the tidal prism for the stage 3 dredge plan during the period between July 27, 1993 and July 27,1994. m<rao TIDAL PRISM PROBABILITY DENSITY FUNCTION DUE TO STAGE 4 DREDGING: 27 JULY 1993 - 27 JULY 1994 6 5-- 3- 1 -- 0 0 10 20 30 40 50 60 70 TIDAL PRISM ( MILLION CUBIC FEET ) 80 FIGURE 74: Incremental probability density function of the tidal prism for the stage 4 dredge plan during the period between July 27, 1993 and July 27, 1994. I • 128 Figures 75-78 show how these increases in the net tidal prism through the four stages of the • dredging plan have led to increases in the tidal velocities U0 at the ocean inlet. A spring-neap cycle from the perigean spring tide epoch of the surrogate year is shown to delineate extreme • maximum and minimum flows during a typically high user demand period in late summer. Each plot shows the existing inlet velocities as a solid line versus the inlet velocities that will occur • during each stage of the dredge plan shown as a dashed line. Of particular significance is how the dredge plan beginning with stage 1 in Figure 75 manages to increase the ebb flow velocities g above the threshold scour velocity of the sediment. This allows for scour and flushing of the recharge zone during ebb flow according to a rate defined by equation (27). With each of the H following stages of the dredge plan, the interval of time during which the ebb flow velocities exceed the threshold velocity progressively increases (see Fiugres 76-78), thereby diminishing | the recharge rate according to equation (17) and hence reducing the closure probability for any given interval of time, At, according to equation (31). For the fully implemented dredge plan • in Figure 78, the ebb flow velocities at the inlet are found to exceed the threshold velocity for _ every ebb tide cycle, even during the neap tides of this high user demand period. This ensures V some degree of inlet scour throughout a high risk situation with combined high plant flow rates I and minimal tidal range. This is the principal mechanism for reducing inlet closure probability. Histograms of the daily closure probabilities for each of the four phases of the dredge I plan are shown in Figures 79-82. Each plot shows the numbers of days in a two-year maintenance dredging cycle that elevated closure probabilities will occur under existing •• conditions (the solid line) versus the conditions of the individual stages of the dredge plan (the • dashed or dotted lines). Beginning with even the first stage of the dredge plan in Figure 79, • there is a decisive reduction in the numbers of days when the highest closure probabilities occur; with reductions in the number of days of closure probabilities of Pc ~ 0.26 to only 68 days for • stage 1 vs. 103 days for existing conditions. For the second stage in Figure 80, these high risk days when closure probabilities of 26% occur are reduced to only 43 in number, and only 17 days for stage 3 (see Figure 81). By stage 4 (Figure 82), the high risk days with 26% closure probabilities are totally eliminated, and no days with closure probabilities in excess of 17% occur at all. Stage 4 has only 38 days out of a two year dredging cycle with closure probabilities of 10%. I I I I o CD OT O O-J INLET VELOCITIES, CHANNEL AXIS AT HYW 101 BRIDGE DREDGE PLAN: STAGE 1, SPRING - NEAP CYCLE 3- -2-- -3 -4 -5 — EXISTING CONDITIONS ••- DREDGE PLAN: STAGE 1 THRESHOLD VELOCITY EBB V 320 340 360 380 400 420 440 460 480 TIME ( HOURS RELATIVE TO 00:00, 1 AUGUST 1993 ) FIGURE 75: Inlet channel velocities for the stage 1 dredge plan vs. existing conditions during a spring-neap cycle with high user demand plant flow rates. o <DW O O INLET VELOCITIES, CHANNEL AXIS AT HYW 101 BRIDGE DREDGE PLAN: STAGE 2, SPRING - NEAP CYCLE -3 -4 -5 EXISTING CONDITIONS DREDGE PLAN: STAGE 2 THRESHOLD VELOCITY EBB V 320 ' 340 360 380 400 420 440 460 480 TIME ( HOURS RELATIVE TO 00:00, 1 AUGUST 1993 ) FIGURE 76: Inlet channel velocities for the stage 2 dredge plan vs. existing conditions during a spring-neap cycle with high user demand plant flow rates. o <DW Oo INLET VELOCITIES, CHANNEL AXIS AT HYW 101 BRIDGE DREDGE PLAN: STAGE 3, SPRING - NEAP CYCLE c FLOOD EXISTING CONDITIONS - DREDGE PLAN: STAGE 3 THRESHOLD VELOCITY 320 340 360 380 400 420 440 460 480 TIME ( HOURS RELATIVE TO 00:00, 1 AUGUST 1993 ) FIGURE 77: Inlet channel velocities for the stage 3 dredge plan vs. existing conditions during a spring-neap cycle with high user demand plant flow rates. o0) CO O O -4 INLET VELOCITIES, CHANNEL AXIS AT HYW 101 BRIDGE DREDGE PLAN: STAGE 4, SPRING - NEAP CYCLE 4 3 2 1 0 -1 •2 -3 -4 -5 FLOOD A EXISTING CONDITIONS DREDGE PLAN: STAGE 4 THRESHOLD VELOCITY 320 340 360 380 400 420 440 460 480 TIME ( HOURS RELATIVE TO 00:00, 1 AUGUST 1993 ) FIGURE 78: Inlet channel velocities for the stage 4 dredge plan vs. existing conditions during a spring-neap cycle with high user demand plant flow rates. 120 CO Q &LH O PQ 100-- 80- 60- 40- 20- 0.0 HISTOGRAM OF CLOSURE PROBABILITIES FOR A TYPICAL BI-ANNUAL DREDGE CYCLE EXISTING CONDITIONS STAGE 1 DREDGE PLAN 0.1 0.2 CLOSURE PROBABILITY 0.3 0.4 FIGURE 79: Histogram of the numbers of days of enhanced closure probability during a biannual dredge cycle for existing conditions vs. the stage 1 dredge plan. 120 100-- co><< Q PtH O KH PQ S !=> 0 HISTOGRAM OF CLOSURE PROBABILITIES FOR A TYPICAL BI-ANNUAL DREDGE CYCLE EXISTING CONDITIONS STAGE 2 DREDGE PLAN 0.1 0.2 0.3 CLOSURE PROBABILITY 0.4 FIGURE 80: Histogram of the numbers of days of enhanced closure probability during a biannual dredge cycle for existing conditions vs. the stage 2 dredge plan. 120 100-- 80-- Q &H O K WPQ S Djz; 20-- 0 HISTOGRAM OF CLOSURE PROBABILITIES FOR A TYPICAL BI-ANNUAL DREDGE CYCLE EXISTING CONDITIONS STAGE 3 DREDGE PLAN 0.0 0.1 0.2 0.3 CLOSURE PROBABILITY 0.4 FIGURE 81: Histogram of the numbers of days of enhanced closure probability during a biannual dredge cycle for existing conditions vs. the stage 3 dredge plan. 120 w Q Pn O OQ 100-- 80-- 60-- 40- 20- HISTOGRAM OF CLOSURE PROBABILITIES FOR A TYPICAL BI-ANNUAL DREDGE CYCLE EXISTING CONDITIONS STAGE 4 DREDGE PLAN 0.1 0.2 0.3 CLOSURE PROBABILITY 0.4 FIGURE 82: Histogram of the numbers of days of enhanced closure probability during a biannual dredge cycle for existing conditions vs. the stage 4 dredge plan. I 137 • The accumulation of risk of closure for each stage of the dredge plan during a bi-annual dredge cycle is calculated from the closure probability distribution function and is plotted in I Figure 83. In Figure 83, the closure probability distribution function for existing conditions is also given as a solid line which grows at 11 % per year, as shown previously in Figure 44. The | stage 1 dredging reduces this closure risk to 7.5 % per year; stage 2 reduces closure risk to 5 % per year; stage 3 reduces closure risk to 4.2% per year; while the fully implemented four-stage | dredge plan reduces the accumulated risk of closure to only 1.3 % per year. At this rate, closure would become more probable than not 38.5 years after completion of the four-stage dredge plan I assuming user demand for power at the levels indicated in Figure 2. Thus, the dredge plan will _ significantly extend the lifespan of the lagoon in the presence of cooling water utilizations greatly • in excess of that originally anticipated by the designer. I To assess the environmental benefits of the four stages of the dredge plan, the tidal hydraulics model was run on each stage at nominal 70% plant activity (570 mgd constant flow Irate) for the extreme spring tide condition shown in Figure 14; the extreme neap condition shown in Figure 15; and the mean tide conditions shown in Figure 16. The resulting waterlines m and east basin water elevations for the stage 1 simulations are found in Appendix VII; stage 2 ' simulations in Appendix VEI; stage 3 in Appendix IX; and stage 4 in Appendix X. A summary m of the results of the hydraulic performance from the simulations given in these appendices appears in Table Vn. Table Vn contrasts the tidal prisms and the subtidal and intertidal areas • for each of these dredge plan stages against the existing conditions and the original construction profile. Inspection of Table 7 reveals that at 70% of maximum plant activity, the fully • implemented four-stage dredge plan recovers nearly all of the original mean tidal prism, and increases the maximum spring tidal prism by nearly 8.25 million cubic feet more than the ft original construction profile. These increases account for the diminished closure probabilities in Figure 82 and the diminished accumulated risk of closure in Figure 83. ft These performance gains were achieved with certain additional gains in habitat values within the lagoon system. For example, the mean intertidal area so critical to birds was • increased 31% over the original construction profile, and increased 23% over the existing conditions. This was achieved by the mix of grading slopes and increases in east basin tidal • range (see Appendix X), which the four-stage dredging plan produced. Furthermore, these gains in the mean intertidal area were not achieved by sacrificing subtidal area in order to preserve I I I I I I I I I I I I I I I I I I I I I 138 mudflats like those left remaining in the east basin (see Figure 69). Instead, the mean subtidal area was slightly increased over the existing conditions (236.9 subtidal acres for stage 4 vs. 233 subtidal acres for existing conditions). There were larger gains in subtidal area for the extreme spring tides which were achieved at the expense of intertidal acreage, but these tradeoffs are only rarely experienced, once every 4.5 years. The intertidal area for stage 4 during the extreme spring tide is still significantly greater than what it was for the original construction profile in 1954 (71.5 intertidal acres for stage 4 vs. 58.1 intertidal acres in 1955). TABLET SUMMARY OF HYDRAULIC PERFORMANCE OF AGUA HEDIONDA DREDGING SEQUENCE (70% Plant Activity, 570 mgd Flow Rate) Spring Tidal Prism Mean Tidal Prism Spring Subtidal Area Mean Subtidal Area Spring Intertidal Area Mean Intertidal Area Original Construction Profile (1954) 80,000,000 55,000,000 235 (Acres) 251 (Acres) 58.1 (Acres) 29.6 (Acres) Existing Conditions November 95 Sounding 63,636,515 (Ft3) 35,541,937 (Ft3) 189.6 (Acres) 233 (Acres) 90.11 (Acres) 31.6 (Acres) Stage 1 Dredging 69,518,492 (Ft3) 42,452,387 (Ft3) 178.3 (Acres) 232 (Acres) 101.6 (Acres) 35.5 (Acres) Stage 2 Dredging 72,412,713 (Ft3) 45,085,563 (Ft3) 176.4 (Acres) 229 (Acres) 106.2 (Acres) 38.3 (Acres) Stage 3 Dredging 76,834,563 (Ft3) 45,715,523 (Ft3) 214.9 (Acres) 236.9 (Acres) 61.8 (Acres) 30.8 (Acres) Stage 4 Dredging 88,249,822 (Ft3) 52,678,459 (Ft3) 222.9 (Acres) 236.9 (Acres) 71.5 (Acres) 38.9 (Acres) COo O w COt— I K Q U O 0.00 CLOSURE PROBABILITY DISTRIBUTION FUNCTION DURING A TYPICAL BI-ANNUAL DREDGE CYCLE EXISTING CONDITIONS STAGE 1 DREDGE PLAN STAGE 2 DREDGE PLAN STAGE 3 DREDGE PLAN STAGE 4 DREDGE PLAN 100 200 300 400 TIME ( DAYS ) 500 600 700 FIGURE 83: Accumulated risk of inlet closure during a typical biannual dredge cycle for existing conditions vs. the four stages of the reconstruction dredge plan. I 140 • vn. CONCLUSIONS A) There is a non-negligible risk of inlet closure for existing conditions. • Under the assumptions that the calendar year of 27 July 93 to 27 July 94 is a statistically representative year, and that the lagoon bathymetry is correctly represented by the • June 1993 soundings, with subsequent updates in November 1995, the following conclusions apply: I 1. During a standard two-year dredge cycle, closure probabilities will exceed I 10% on 361 days out of 730 days, or 49% of the time. 2. The most common episode of elevated closure probability is on the order gj of 25% and involves the independent occurrence of any one of the following factors: 1) high waves, 2) oblique wave directions, 3) neap | tides, or 4) high plant inflow rates (greater than 600 mgd). Independent occurrence of any of these aggravating factors happens about 103 days | throughout a standard two-year dredge cycle. The simultaneous _ occurrence of more than one of these aggravating factors is generally rare 8 in a typical year, occurring only once between 27 July 93 and 27 July 94 — and resulting in a closure probability of 38%. • 3. The accumulated risk of inlet closure under normal plant operating conditions grows at about 11% per year. Consequently, inlet closure inI I I I I I more probable than not after 4.5 years. B) Three remedial approaches to reduce inlet closure risk were studied: 1) m manipulation of plant flow rates, 2) modification of choke points, and 3) reconstruction * dredging. Dredging and flow rate modifications were found to be effective, but dredging • is the most compatible solution with existing plant operations. 1. Manipulation of Plait Flow Rates: Manipulation of plant inflow rates can dramatically reduce both closure probability and accumulated risk of inlet closure. A schedule of plant inflow rate reductions or increases in response to the occurrence of I I I I I I I I I I I I I I I I I I 141 aggravating factors listed in Conclusion A-2 above can reduce most inlet closure probabilities to less than 3 %, while reducing accumulated risk of inlet closure to 1V* % per year. Manipulation of plant inflow rates does not appear to be a practical alternative to routine maintenance dredging. 180 separate manipulations of plant inflow rates would be necessary in a typical calendar year to achieve these benefits. More than half of these manipulations would involve deep cuts in plant inflow rates to levels between 13 and 163 mgd. All manipulations must be coordinated with ebbing tide. Manipulations of inflow rates might, however, remain a viable option for a short term emergency if the aggravating factors listed in Conclusion A-2 above can be anticipated sufficiently far in advance. 2. Choke Point Modification: Two separate schemes for modifying the choke points to reduce head losses and recover tidal prism were studied, namely the "Stratford Fence," which uses a flat bottom and the "Stratford Bottom," which uses a variable depth bottom. The flat bottom "Stratford Fence" is slightly better and involves less rock or ArmorFlex work. Mean tidal range losses were reduced from 1.80 ft to only 0.778 ft for the "Stratford Fence," and 0.588 ft for the variable depth "Stratford Bottom". As a result, 15.4% of the mean tidal prism was recovered by the "Stratford Fence" concept, while 11.9% was recovered by the "Stratford Bottom" concept. The benefits derived from these recoveries of tidal prism were reductions in the accumulated risk of closure to 5% per year for the "Stratford Fence" and 6% per year for the "Stratford Bottom". The choke point modification schemes are therefore less effective than the other alternatives studied and are also expensive to build, with cost estimates ranging between $1.3 and $1.5 million. 3. Reconstruction Dredging: A dredging plan with a four-stage production schedule was studied. When fully implemented the dredge plan will recover most of the mean tidal I • prism of the original construction profile completed in 1954, yielding 52.7 million cubic ft of mean prism vs. 55 million cubic ft for the original • post-construction lagoon system. The dredging plan will actually increase the spring tidal prism to 88.25 million cubic ft as compared to 80 million • cubic ft for the original construction profile. The dredging plan will also increase the mean intertidal area to 38.9 acres vs. 31.6 for the 1954 post-I construction lagoon. Mean subtidal area will also be increased slightly to 236.9 acres vs. 233 acres for the existing lagoon. In addition to these P increases in available habitat, the dredge plan will also remove most of the fine sediments which are resuspended by boating activities and daily wind • mixing. Therefore the dredge plan will improve water visibility, which _ should promote the recruitment of eel grass and increase the overall I biomass and productivity of the lagoon in general. The fully implemented _ dredge plan is expected to reduce instantaneous closure probabilities to no V more than 12% for 20 days each year, and reduce the accumulated risk of closure to only 1.3 % per year. Thus the closure interval for the four-I I I I I I I I I I stage dredge plan will be extended to 38 years. I REFERENCES I Bagnold, R. A. , 1962, " Autosuspension of transported sediment; turbidity currents, " Proceedings of the Royal Society of London, Series A, v.265, p. 315-319. I Bagnold, R.A., 1963, "Mechanics of marine sedimentation in the sea," M.N. Hill, ed., Interscience, v. 3, p. 507-528. | Bagnold, R.A., 1966, "An approach to the sediment transport problem from general physics," 17.5. Geological Survey, Professional Paper 422-1, 37 pp. V Boas, Mary L. , 1966, Mathematical Methods in the Physical Sciences, John Wiley & Sons, New York, 778 pp. I Coastal Data Information Program, 1993-1994, "Monthly Summary Reports," SIO Reference Series 93-27 through 94-19. • Ellis, J.D. , 1954, "Dredging Final Report, Agua Hedionda Slough Encina Generating Station, " San Diego Gas and Electric, 44 pp. • Flick, R.E., 1991, "Joint occurrence of high tide and storm surge in California," Proc. World Marina '91, First Int'l Conf.t ASCE, p. 60-62. • Flick, R.E., and Badan-Dangon, A., 1989, "Coastal sea levels during the January 1988 Storm off the Californias," Shore and Beach, v.57, n.4, p. 28-31. Flick, R.E., and Cayan, D.C., 1984, "Extreme sea levels on the coast of California," Proc. 19th _ Int'l Conf. Coastal Eng., American Soc. Civil Eng., p. 886-898. Inman, D.L., and S.A. Jenkins, 1985, "Erosion and accretion waves from Oceanside harbor," I pp. 591-93 in Oceans 85: Ocean Engineering and the Environment, Marine Technological Society and TPMR v. 1, 674 pp. I Inman, D.L., and S.A. Jenkins, 1983, "Oceanographic report for Oceanside Beach facilities," Report to the City of Oceanside, California, unpublished, 206 pp. I Jenkins, S.A., and J. Wasyl, 1994, "Numerical modeling of the tidal hydraulics and inlet closures at Agua Hedionda Lagoon, Part n: Risk Analysis," submitted to SDG&E, 60 pp. I Jenkins, S.A., and J. Wasyl, 1993, "Numerical modeling of tidal hydraulics and inlet closures at Agua Hedionda Lagoon", submitted to SDG&E, 91 pp. • Jenkins, S.A., and D.W. Skelly, 1989, "An evaluation of the coastal data base pertaining to seawater diversions at Encina power plant", SIO Reference Series, #89-4, 52 pp. I I 144 • Jenkins, S.A., Skelly, D.W., and J. Wasyl, 1989, "Dispersion and momentum flux study of the cooling water outfall at Agua Hedionda," SIO Reference Series, #89-17, 36 pp. • Kenlegan, G.H. and W.C. Krumbein, 1949, "Stable configurations of bottom slope in a shallow sea and its bearing on geological processes," Trans. American Geophysical Union, v. 30, _ n. 6, p. 855-861. Komar, P.D. and D.L. Inman, 1970, "Longshore sand transport on beaches," Jour. Geophysical • Res., v. 75, n. 30, p. 5914-5927. Kraus, N.C. and S. Harikai, 1983, "Numerical model of the shoreline change at Onai Beach," • Coastal Engineering, v. 7, n. 1, p. 1-28. Larson, Harold J., 1969, Introduction to Probability Theory and Statistical Inference, John • Wiley & Sons, New York, 387 pp. Leighton and Associates, 1988, "Report of test results and laboratory analysis, San Diego Gas I and Electric Encina Plant Dredge Material, Carlsbad, California: Project No. 4860085- 02." I Liebeck, R. H., 1976, "On the design of subsonic airfoils of high lift," paper no. 6463, McDonnell Douglas Tech. Report, 25 pp. I Liebeck, R.H., and Ormsbee, A.I., "Optimization of Airfoils for Maximum Lift," AIAA Journal of Aircraft, v.7, n. 5, Sept-Oct 1970. • McCormick, B., 1979, Aerodynamics, Aeronautics and Flight Mechanics, John Wiley & Sons,m New York, 652 pp. I National Oceanic and Atmospheric Administration, May 1972, "Gulf of Santa Catalina," Bathymetry Chart, No. 18774. • Nearshore Research Group, 1984, "Nearshore Bathymetric Survey Report," No. 1, CCSTWS 84-2. | O'Brien, M.P, 1931, "Estuary Tidal Prism related to Entrance Areas," The Journal, Civil Engineering, v. 8, n. 1, p. 738-739. | Ozasa, H., and A. H. Brampton, 1980, "Mathematical modeling of beaches backed by walls," Coastal Engineering, v. 4, n. 1, p. 47-64. I Smith, A.M.O., High-Lift Aerodynamics, Wright Brothers Lecture, AIAA Paper No. 74-939m August 1974. • Stratford, B.S., The Prediction of Separation of the Turbulent Boundary Layer, Jour. Fluid Mech.,v. 5, 1959. I I 145 U.S. Coast and Geodetic Survey, Air Photo Compilation No. T-5412, January 17,1934, 10:30 a.m. I Stratford, B.S., An Experimental Flow with Zero Skin Friction Throughout its Region of ™ Pressure Rise, Jour. Fluid Mech., v. 5, 1959. I Sucsy, P.V., Pearce, A.M., and V.G. Panchang, 1989, "Sensitivity of Gulf of Maine tide model to depth," Estuarine and Coastal Modeling, ASCE, New York, pp. 268-277 I U.S. Army Corps of Engineers, 1985, Shore Protection Manual. I U.S. Army Corps of Engineers, Los Angeles District, 1987, "Oceanside Littoral Cell Preliminary Sediment Budget Report," CCSTWS 87-4, 158 pp. • U.S. Army Engineer District, Los Angeles, Corps of Engineers, 1970, San Diego County, Beach Erosion Control Repor of Coast of Southern California, Three-Year Report, 1967- 1968-1969. I I U.S. Coast and Geodetic Survey, Hydrographic Survey No. 5648, March-April, 1934. • U.S. Coast and Geodetic Survey, Hydrographic Survey No. 5663, March-May, 1934. USGS, 1976, Final Environmental Statement, Oil and Gas Development in the Santa Barbara • Channel, Department of the Interior, 3 vols. Wortmann, F.X., The Quest for High Lift, AIAA Paper No. 74-1018, September 1974. I I I I I I I I I • APPENDIX I SOURCE CODE TO OCEANRDS_NESTED_GRID • OF REFRACTION/DIFFRACTION MODEL I I I I I I I I I I I I I I I I I I I I I I I I I I I I B I I I I I £************** ********************************************************* c c Ocean refraction - diffraction module oceanrds_nested_grid.f c written by Scott A. Jenkins & Joseph Wasyl c revised 1993,1994,1995,1996 c £*********************************************************************** c dimension arrays to size of bathymetry grid (ni,nj) c parameter(ni=996, nj=996) parameter (max=10000) c character name*8,ifile*12 character ofilel*12,ofile2*12,ofile3*12,ofile4*12 character ofile5*12,ofile6*12,ofile8*12,ofile9*12 character ofilell*12,ofilel3*12 character*17 north,south,east,west,rows,cols,x_res,y_res dimension ccg(max),di(max),hab(max),rib(max),dib(max),ih(max) dimension iyb(ni),bh(ni),ba(ni),rxb(ni) dimension wnum(ni,nj),wht(ni,nj),ang(ni,nj) dimension wnum5x5(200,200),wht5x5(200,200),ang5x5(200,200) dimension depth(ni,nj),depthold(ni,nj),depth5x5(200,200) real kbar(2,max),kave c complex aa(max),bb(max),cc(max),dd(max),uu(max),aprev(max) complex c3,tl,t2,t3,f,alast(max),aphys(max),mim,mip c common /grid/ ny,sy,sx,dely,delx,ndely,freq,dcon,ify,ifx, tide common /cut/ dc c pi=acos(-1.) c c open parameter input file, return error message if missing c c c open(2,file='oceanrds.inp',status='old') read(2,'(a)') name read(2,*) icoast read(2,*) tide_elev read(2,*) sx read(2,*) sy read(2,*) nduml read(2,*) ndum2 read(2,*) persw read(2,*) asw read(2,*) hsw c tide=tide_elev amp=hsw per=persw c freq=l/per c c... set some variables to constant values c c dcon = 1. for nuwave version dcon=l. c c cutoff depth (dc) -5.0 meters dc=-5.0 c c breaking wave switch (ibreak) (l=on, 0=off) ibreak=l I I I I I 8 I I I I I I I I I c breaking criteria (be) 0.5 for wave height = (bc)*depth bc=0.5 c c lateral b.c.: ibc=(0) transmitting, ibc=(l) reflective ibc=0 c if trans: (isn=0) straight snell, isn=(l) kirby's improved isn=l c idf=(0) small angle diff, idf=(l) large angle dif idf=l c c write(*,'(5(/),10x,a)') &' OCEANRDS ' write(*,'(/,10x,a)') &'- based on the parabolic equation method (PEM) of solving' write{*,'(10x,a,5(/))') &' the mild-slope equation. ' c nn=8 do 3 m=8,l,-l 3 if(name(m:m).eq.' ') nn=m-l c c c... open grid file open(20,file='bathy.dat',status='old') c 1953 format(a!7) read(20,1953) north write(*,*) north c read(20,1953) south write(*,*) south c read(20,1953) east write(*,*) east c read(20,1953) west write(*,*) west c read(20,1953) rows write(*,*) rows c read(20,1953) cols write(*,*) cols c read(20,1953) x_res write(*,*) x_res c read(20,1953) y_res write(*,*) y_res c read(20,*) coast write(*,*) coast dir=asw c c... change to proper rotation frame, flag if theta not between + /- 45 c icoast: l=west; 2=north; 3=east; 4=south theta=coast-dir c c c*********breaker file for pvm 200x200 50 meter grid c... open breaker ix,iy,wave height, wave angle, depth file write(ofile9,'(a,a)') namet:nn),'.brl' open(49,file=ofile9,status='unknown') c I I I I I I I I I I I I I I c c... open rotated ascii bathymetry file for internal computation write(ofilei,'(a,a)') name(:nn),'.dep' open(31,file=ofilel,status='unknown') c c c... open pvm grid bathymetry file 200x200m: every 5 cells sampled write(ofilell,'(a,a)') name(:nn),'.grd' open(41,file=ofilell,status='unknown') c c... open ascii wave number file write(ofile2,'(a,a)') name(:nn),'.wvn' open(32,file=ofile2,status='unknown') c c... open ascii wave height file write(ofile3,'(a,a)') name(:nn),'.whl' open(33,file=ofile3,status='unknown') c c... open ascii wave angle file write(ofile4,'(a,a)') name(:nn),'.anl' open(34,file=ofile4,status='unknown') c c... open ascii wave height file write(ofile8,'(a,a)') name(:nn),'.wht' open(35,file=ofile8,status='unknown') c write(ofile!3,'(a,a)') name(:nn),'.swl' open(25,file=ofile!3,status='unknown') c open(24,file='break.dat',status='unknown') c c c read depth array raster format ( upper left corner:i=l,j=l ) c ilijl must be lower left corner of deepwater offshore boundary c i goes from deep to shallow c c icoast= l=west; 2=north; 3=east; 4=south ^ikik^^Hfilf^lt^r^t^ifrifr^«Qag^ ^O131 lOTl ^^^^f^**^^^^ if(icoast.EQ.l)then do 111 j=nj,l,-l read(20,*) (depthold(i,j),i=l,ni) 111 continue rewind(20) endif c if(icoast.EQ.2)then do 212 i=l,ni read(20,*) (depthold(i,j),j=l,nj) 212 continue rewind(20) endif c if(icoast.EQ.3)then do 313 j=l,nj read(20,*) (depthold(i,j),i=ni,1,-1) 313 continue rewind(20) endif if(icoast.EQ.4)then do 414 i=ni,l,-l read(20,*) (depthold(i,j),j=nj,l,-l) 414 continue rewind(20) endif I I I I I I I I 1 I c c....correct depth for tide elevation do 728 ii=l,ni do 729 jj=l,nj depthold(ii,jj)=depthold(ii,jj)+tide_elev 729 continue 728 continue c cccccccccccccc j goes parallel to shore i goes offshore to onshore c 165 format(996f7.2) c open(100, file='5x5test.dat',status='unknown') c if(persw.LE.6.0)then nx=ni ny=nj c c..write left justified computational lOxlOm file deepwater column to shore DO 101 1=1,ni WRITE(31,165)(depthold(I,J),J=l,nj) 101 CONTINUE rewind(31) endif c if(persw.GT.6.0)then nx=200 ny=200 sx=5.0*sx sy=5.0*sy newi=0 icount=5 c do 102 i=l,ni if(icount.EQ.5)then icount=0 newi=newi+l jcount=5 newj=0 do 103 j=l,nj if(jcount.EQ.5)then jcount=0 newj=newj+l depth5x5(newi,newj)=depthold(i,j) endif jcount=jcount+l 103 continue endif icount=icount+l 102 continue c c..write right justified computational file DO 104 1=1,nx WRITE(31,166)(depth5x5(I,J),J=l,ny) 104 CONTINUE rewind(31) endif c if(persw.LE.6)then do 801 i=l,ni read(31,*) {depth(i,j),j=l,nj) 801 continue else do 811 i=l,nx read(31,*) (depth(i,j),j=l,ny) 1V 1 1 1 1 t 1•w 1 1 i i i 811 continue endif rewind (31) €&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&&& C c. . C c. . c c c c. . c c c . read first depth to initialize offshore boundary read{31,*) (di (m) ,m=l,ny) ridep=di (ny/2) rewind (31) . . would like a grid step size on the order of 1/5 wavelength call getkcgdO. ,wk,cgO) wl=2*pi/wk ifx=l+sx/(.2*wl) ify=l+sy/(.2*wl) nmx=ifx nmy=ify amp=amp/2 . . calculate dimensions of interpolated grid dely=sy/ify delx=sx/ifx ndely= (ny-1) *ify ndelx= (nx-1) *ifx call getkcg(ridep,wkO,cgO) c wave length (m) wl=2*pi/wkO c wave frequency (rad/sec) c 33 c c c. . c c. . c c. . 555 c c sig=2*pi/per radder's correction factor do 33 j=l,ndely+l ccg( j)=sqrt (sig*cgO/wkO) ltype=2 . open binary wave height file write (ofileS, ' (a, a) ' ) name( :nn) , ' .bwl' open(9, f ile=of ile5, form=' unformatted' ) . open binary wave angle file write (of ile6, ' (a, a) ' ) name( :nn) , ' .bal' open (11, file=ofile6, form=' unformatted' ) initialize depth array do 555 nn=l,ndely+l hab(nn) =0. ih(nn)=0 continue enter initial condition at x=0 c call inbc(amp.wkO,theta,ndely,dely,alast) c do 202 j=l,ndely+l kbar(1,j)=wk 202 kbar(2,j)=wk c c scale alast as in radder(1978) I I I I I I I I I I I I I I do 32 j=l,ndely+l alast(j)=alast(j)*ccg(j) 32 continue c c call wwave(alast,ndely,nmy,dely,kbar) c c solution of the parabolic eqn. by the crank-nicholson formulation c c start x increments c c2=l./2./dely**2 c3=2.*(0,l)/delx c increments in x-direction c ikount=0 do 100 l=l,ndelx c c... write every 10th step to the screen c if(10*((1+1)710).eq.1+1) then print *,' column ',1+1,' of ',ndelx+l endif c lm=l c call intkcgdm, kbar,ccg,di, ikount) c c correction factor do 34 j=l,ndely+l ccg(j)=sqrt(sig*ccg(j)/kbar(2,j)) 34 continue c c increments in y-direction - 1st. round do 200 j=2,ndely c kave=(kbar(2,j)+kbar(l,j))12. c kave=kbar{2,j) c if(idf.eq.O) then c small angle diffraction approximation c tl=c2 t3=c3*kave f=2*kave*(kave-wk)+c3/2*(kbar(2,j)-kbar(l,j)) c aa(j-l)=tl bb(j-l)=t3-2.*tl+f/2. cc(j-l)=tl dd(j-l)=-tl*alast(j+l)+(t3+2*tl-f/2)*alast(j)-tl*alast(j-l) c endif c if(idf.eq.l) then c large angle diffraction approximation c tl=c2/2.*(3.-wk/kave) t2=c2*c3/4./kave t3=c3*kave f=2*kave*(kave-wk)+c3/2*(kbar(2,j)-kbar(l, j)) c aa(j-l)=tl+t2 bb(j-l)=t3-2.*(tl+t2)+f/2. cc(j-l)=tl+t2 I I I 1 1 t I I t I I I I I I dd(j-l)=(t2-tl)* (alast (j+l)+alast(j-l) ) + (t3-2* (t2-tl) - & *alast(j) endif c if(j.eq.2) then c if(ibc.eq.l) cc (1) =aa (1) +cc (1) c reflective b.c. : al=a3 if(ibc.eq.l) bb(l) =aa (1) +bb(l) c reflective b.c. : al=a2 if(ibc.eq.O) then c transmitting b.c. c if(isn.eq.O) then c straight snell mip= (0. ,1. )*wk*sin(theta) /2. mip= ( 1 . /dely-mip) / ( 1 . /dely+mip) else c kirby ' s improved mip= (alast (2) -alast (1) ) / (alast (2) +alast (1) ) mip= ( 1 . -mip ) / ( 1 . +mip ) endif c bb(l)=bb(l)+mip*aa(l) endif endif c if ( j .eq.ndely) then c if(ibc.eq.l) aa(j-l) =aa (j-1) +cc(j-l) c reflective b.c. : an=an-2 if (ibc.eq.l) bb(j-l) =bb(j-l) +cc(j-l) c reflective b.c. : an=an-l if(ibc.eq.O) then c transmitting b.c. c if(isn.eq.O) then c straight snell mim= (0. ,l.)*wk*sin(theta) /2. mim= ( 1 . /dely+mim) / ( 1 . /dely-mim) else c kirby ' s improved mim= (alast ( j+1) -alast ( j ) ) / (alast ( j+1) +alast ( j ) ) mim= ( 1 . +mim) / ( 1 . -mim) endif c bb ( j -1 ) =bb ( j -1 ) -t-mim*cc ( j -1 ) endif cc(j-l)=<0.,0.) endif c 200 continue c c c ***** compute the solution ***** c neqs=ndely-l c call tridag(aa,bb,cc,dd,uu,neqs) c do 143 j=2,ndely 143 alast (j)=uu( j-1) c load alast ( j) c if (ibc.eq.l) thenI t I I I I I I I I I I I 1 I I I c alast(l)=alast(3) c reflective b.c.: al=a3 c alast (ndely+1) =alast (ndely-1) c : an=an-2 alast (l)=alast (2) c reflective b.c.: al=a2 alast (ndely+1 ) =alast (ndely ) c : an=an-l endif c if(ibc.eq.O) then c transmitting b.c. c alast (1)= mip*alast(2) alast (ndely+1) = mim*alast (ndely) c endif c c transform back into phys. height do 37 j=l, ndely+1 37 aphys (j)=2.*alast (j) /ccg(j) c c ********* check for breaking ********* c if (ibreak.eq. 1) then do 54 j=l, ndely+1 hb=bc*di(j) rat=hb/cabs (aphys ( j ) ) if (di(j) .lt.0) rat=0. if (rat. It. 1) then c save point just before wave first breaks if (ih(j) .eq.O) then hab ( j ) =cabs ( aprev ( j ) ) rlb(j)=(lm-l)/real(ifx) c calculate direction before breaking if ( j . eq. ndely+1. or. ih( j+1) .eq.l) then xx=aimag ( (aprev( j ) -aprev( j-1) ) / (dely*kbar (1, j ) *aprev( j ) ) ) if(xx.gt.l) xx=l. if (xx.lt. -1.) xx=-l. dib( j ) =asin(xx) else xx=aimag( (aprev( j+1) -aprev( j) ) / (dely*kbar (1, j ) *aprev( j ) ) ) if(xx.gt.l) xx=l. if (xx. It . -1. ) xx=-l. dib( j ) =asin(xx) endif dib ( j ) =270 . -dib ( j ) * 57 . 296-rot if (dib(j) .lt.0) dib(j)=360+dib(j) endif aphys ( j ) =rat*aphys ( j ) alast ( j ) =rat*alast ( j ) endif 54 continue endif c c c... writing wave field if (nmx* (Im/nmx) . eq. 1m. or. 1m. eq.l) then call wwave ( aphys , ndely , nmy , dely , kbar ) endif c do 55 nn=l, ndely+1 aprev ( nn ) =aphy s ( nn ) I I I I I I I I i i i 55 continue c 100 continue c c c... output prebreak heights and directions m=l write(24,'(122,f12.2,f12.1,f12.1)') m,hab(m),rlb(m),dib(m) alphab=dib(m)-shor irlb=NINT(rlb(m))+l breakd=(5.0/4.0)*hab(m) c write(39,'(18x,14,14,f12.3,f12.1,f!2.3)') m,irlb.hab(m), c &alphab,breakd c. . if(nmy.eq.l) then is=2 else is=nmy endif mm=l do 888 m=is,ndely,nmy mm=mm+l write(24,'(122,f12.2,f12.1,f12.1)') mm,hab(m),rlb(m),dib(m) alphab=dib(m)-shor irlb=NINT(rlb(m))+l breakd=(5.0/4.0)*hab(m) c write(39,'(18x,14,14,f12.3,f12.1,f12.3)')nun,irlb,hab(m), c &alphab,breakd 888 • continue c rewind(24) c if(persw.LE.6)then do 150 n=l,ni read(24,*) iyb(n),bh(n),rxb(n),ba(n) 150 continue close(24) ncount=5 newn=0 do 770 n=l,ni i f(ncount.EQ.5)then ncount=0 newn=newn+l if(icoast.EQ.2)ba(n)=ba(n)-180.0 if(icoast.EQ.3)ba(n)=ba(n)+90.0 if(icoast.EQ.4)ba(n)=ba(n)-90.0 write(25,'(18x,14,f12.2,f!2.1,f12.1)')newn,bh(n),rxb(n),ba(n) endif ncount=ncount+l 770 continue write(*,*)newn else do 152 n=l,ny read(24,*) iyb(n),bh(n),rxb(n),ba(n) 152 continue close(24) do 151 n=l,ny if(icoast.EQ.2)ba(n)=ba(n)-180.0 if(icoast.EQ.3)ba(n)=ba(n)+90.0 if(icoast.EQ.4)ba(n)=ba(n)-90.0 write(25,'(18x,14,f12.2,f12.1,f12.1)')iyb(n),bh(n),rxb(n),ba(n) 151 continue endif c call disp(per,depth,pi,amp,wnum,ni,nj,nx,ny) I I I I I I I I I I I I I I c rewind (9) rewind (11) c c read wave height and wave angle arrays into program do 114 i=l,nx read(9) (wht(i,j),j=l,ny) read(ll) (ang(i,j),j=l,ny) 114 continue c c 960 continue 940 continue 950 CONTINUE 930 CONTINUE c.... write array values to: c.... ascii depth, wave number, wave angle, wave height grid files c raster format c c**********************iCOAST ***************************** c if(persw.LE.6.0)then nx=200 ny=200 icount=5 newi=0 c do 702 i=l,ni i f(icount.EQ.5)then icount=0 newi=newi+l jcount=5 newj=0 do 703 j=l,nj i f (j count.EQ.5)then jcount=0 newj=newj+l depthSxS(newi,newj)=depthold(i,j) wnum5x5(newi,newj)=wnum(i,j) wht5x5(newi,newj)=wht(i,j) ang5x5(newi,newj)=ang(i,j) endif jcount=jcount+l 703 continue endif icount=icount+l 702 continue c c..write 200x200 files - raster format if(icoast.EQ.l)then DO 761 J=ny,l,-l WRITE(41,166) (depth5x5(i,j),i=l,nx) write(32,166) (wnum5x5(i,j),i=l,nx) write(33,166) (wht5x5(i,j),i=l,nx) write(34,166) (ang5x5(i,j),i=l,nx) write(35,166) (wht5x5(i,j),i=l,nx) 761 CONTINUE endif c if(icoast.EQ.2)then DO 762 i=l,nx WRITE(41,166)(depth5x5(i,j),j=l,ny) write(32,166) (wnum5x5(i,j),j=l,ny) write(33,166) (wht5x5(i,j),j=l,ny) write(34,166) (ang5x5(i,j),j=l,ny) 1 1wi 1"W I 1 • 1•• 1 1 1 1 1 1 1 I 1 write(35,166) (wht5x5(i, 762 CONTINUE endif c c if (icoast .EQ.3) then DO 763 j=l,ny WRITE (41, 166) (depthSxS (i, write(32, 166) (wnum5x5(i write(33,166) (wht5x5(i, write(34, 166) (ang5x5(i, write(35,166) (wht5x5(i, 763 CONTINUE endif c if ( icoast . EQ . 4 ) then DO 764 i=nx,l,-l WRITE (41, 166) (depthSxS (i, write (32, 166) (wnum5x5(i write (33, 166) (wht5x5(i, write(34, 166) (ang5x5(i, write (35, 166) (wht5x5(i, 764 CONTINUE endif c endif c c 166 FORMAT(200f8.3) c if (persw.GT. 6.0) then nx=200 ny=200 c c.. write 200x200 PVM 50x50m files if ( icoast. EQ.l) then DO 661 J=ny,l,-l WRITE (41, 166) (depthSxS (i write (32, 166) (wnum(i,j) write(33,166) (wht(i,j), write(34, 166) (ang(i,j), write (3 5, 166) (wht(i,j), 661 CONTINUE endif c i f ( icoast . EQ . 2 ) then DO 662 i=l,nx WRITE(41,166) (depthSxS (i, write (32, 166) (wnum(i,j) write (33, 166) (wht(i,j), write(34, 166) (ang(i,j), write(35,166) (wht(i,j), 662 CONTINUE endif c if (icoast. EQ.3) then DO 663 j=l,ny WRITE(41,166) (depthSxS (i, write(32, 166) (wnum(i,j) write (33, 166) (wht(i.j), write(34,166) (ang(i,j), write(35,166) (wht(i.j), 663 CONTINUE endif c j) , j=l,ny) j) ,i=nx,l,-l) , j ) , i=nx, 1, -1) j) ,i=nx,l,-l) j) , i=nx, 1, -1) j) ,i=nx,l,-l) j) , j=ny,l,-l) . j) . D=ny, 1, -1) j) , j=ny,l,-l) j) , j=ny,l,-l) j) , j=ny,l,-l) - raster format , j) ,i=l,nx) ,i=l,nx) i=l,nx) i=l,nx) i=l,nx) j) , j=l,ny) , j=l,ny) j=l,ny) j=l,ny) j=l,ny) j) ,i=nx,l,-l) , i=nx, 1, -1) i=nx, 1, -1) i=nx, 1, -1) i=nx, 1, -1) I I I I I I I I I I I 1 I I if(icoast.EQ.4)then DO 664 i=nx,l,-l WRITE(41,166)(depth5x5(i,j),j=ny,l,- write(32,166) (wnum(i,j),j=ny,!,-!) write(33,166) (wht(i,j),j=ny,!,-!) write(34,166) (ang(i,j),j=ny,l,-l) write(35,166) (wht(i,j),j=ny,l,-l) 664 CONTINUE endif c endif c c go to 901 900 write(*,'(20x,a,a,a)') '**** error **** ',ifile,' missing' write(*,'(20x,a)') ' press any key to continue ' read(*,'(a)') idum 901 continue stop end c c c subroutine tridag(a,b,c,r,u,n) c c parameter (nmax=10000) complex gam(nmax),a(n),b(n),c(n),r(n),u(n),bet c bet=b(l) u(l)=r(l)/bet do 11 j=2,n gam(j)=c(j-l)/bet bet=b(j)-a(j)*gam(j) if(bet.eq.0)pause u(j)=(r( 11 continue do 12 j=n-l,l,-l u(j)=u(j)-gam(j+1)*u(j+1) 12 continue return end c subroutine wwave(aphys,ndely,nmy,dely,kbar) c c compute & write wave height and direction, initial breaking point c imax= max grid size nmax=max no. of steps c parameter (nmax=10000,imax=10000) dimension amod(nmax),alfa(nmax) real kbar(2,nmax) complex aphys(nmax) c nky=0 do 350 j=l,ndely+l if(nmy*(j/nmy).eq.j.or.j.eq.l) then nky=nky+l c amod(nky)=cabs(aphys(j)) if(amod(nky).gt..05) then if(j.eq.ndely+1) then xx=aimag((aphys(j)-aphys(j-1))/(dely*kbar(2,j)*aphys(j))) 1 1 1 1 1 ^•r 1 1 i I l i 350 c c c c c c c c c 2 c c wa 3 c c cccc c c c. . . c. . . c c if(xx.gt.l) xx=l. if (xx.lt. -1. ) xx=-l. alfa (nky) =asin (xx) else xx=aimag( (aphys ( j+1) -aphys ( j ) ) / (dely*kbar (2, j ) *aphys ( j ) ) ) if(xx.gt.l) xx=l. if (xx.lt. -1. ) xx=-l. alfa (nky) =asin(xx) end if else alfa(nky)=0 endif alfa (nky) =alfa (nky) *57. 296 endif continue write (9) (amod( j j ) , j j=l,nky) write(ll) (alfa(jj) , jj=l,nky) return end subroutine inbc (amp, wk, theta,ndely,dely, aa) complex aa ( * ) real ky pi=acos (-1. ) theta=theta*pi/180 . ky=wk* sin ( theta ) do 3 j=l,ndely+l ar=amp*cos (ky* ( j-1) *dely) ai=amp*sin(ky* (j-1) *dely) ar= (cos (ky* ( j-1) *dely) +cos (-ky* ( j-1) *dely) ) /2 . identical ai=(sin(ky*(j-l)*dely)+sin(-ky*(j-l)*dely))/2. ves summed aa ( j ) =cmplx ( ar , ai ) continue return end ccccccccccccccccccccccccccccccccccccccccccccccccccccccccccccccccc subrout ine getkcg ( d , k , eg ) depth (meters) and frequency are input returns wave number (2pi/l) and group velocity real k common /grid/ ny, sy, sx,dely,delx,ndely, f ,dcon, ify, ifx, tide data tpi/6.2831853/ sig=tpi*f a=d*sig*sig/9.81 if(a.ge.l) then yhat=a*(l+1.26*exp((-1.84)*a)) t=exp( (-2)*yhat) aka=a*(l+2*t*(l+t)) else aka=sqrt(a)*(l+a/6.*(l+a/5.) ) I I I I I I I I I i I I I 1 i i endif k=aka/d x=2*k*d cg=tpi* (f/k)*.5* (l+x/sinh(x) return end c £***********************************! c subroutine intkcg (1m, kbar,ccg,di, ikount) c parameter (max=10000) real kbar(2,max) dimension ccg(max) ,d(2,max) ,di(max) c save d common /cut/ dc common /grid/ ny, sy, sx,dely,delx,ndely, f ,dcon, ify, ifx, tide data istart,ncol,eps/l, 0, .0000001/ c c... if this is the first call, load first two columns c go to (1,3) istart 1 istart=2 c... read in first 2 columns do 777 i=l,2 777 read(31,*) (d(i, j ) , j=l,ny) ncol=l c... interpolate grid points for first row do 2 i=l,ndely y=l+real (i-1) /real (ify) +eps ym=amod(y , 1 . ) iy=int (y) di(i)=d(l,iy)+ym*(d(l,iy+l)-d(l,iy))+tide call getkcg (di (i) *dcon,rk,cg) ccg(i)=cg 2 kbar(2,i)=rk c c... figure out which bathymetry grid columns should be used c 3 icol=l+real (1m) /real (ifx) +eps if (icol.ne.ncol) then do 5 i=l,ny 5 d(l,i)=d(2,i) read(31,*,end=999) (d{2, j ) , j=l, ny) 999 ncol=icol endif ddl=d(l,100) c c... bilinear interpolate depths for new column of k and eg c 15 x=l+real (1m) /real (ifx) +eps xm=amod ( x , 1 . ) do 20 i=l,ndely y=l+real (i-1) /real (ify) +eps ym=amod(y, 1 . ) iy=int(y) & xm*ym*(d(2,iy+l)-d(l,iy+l)-d(2,iy)+d(l,iy) )+tide if (di(i) .le.Ol) then di(i)=.01 endif c c... shift k and ccg col 2 to 1 and calculate new 2's c 1 J 20 1 c £***** 1 C 1 1 c 1IV 1 1 1w 1030 I 1050 1040 109011020 1010 c 1 1 1 1 1 1 if (di(i) .lt.dc) di(i)=dc kbar(l,i)=kbar(2,i) call getkcg(di (i) *dcon, rk,cg) kbar (2, i) =rk ccg(i) =cg kbar ( 1 , ndely+1 ) =kbar ( 1 , ndely ) kbar ( 2 , ndely + 1 ) =kbar ( 2 , ndely ) ccg (ndely+1) =ccg (ndely) di (ndely+1) =di (ndely) return end subroutine disp (per, depth, pi, amp, wnum,ni,nj ,nx,ny) dimension depth (ni, nj ), wnum(ni,nj) freq=l/per sigma=2 . 0*pi*freq amp2=amp*2 g=9.8 do 1010 i=l,nx do 1020 j=l,ny initialize wave number array wnum(i, j ) =-1. 0 if (depth(i, j) .LE.O)go to 1090 y= ( sigma* *2 *depth ( i , j ) ) /g if (depth(i, j ) . LT . amp2 ) then wnum(i, j ) =sigma/ (g*depth(i, j ) ) **0. 5 go to 1090 else endif x=y do 1030 k5=l,100 h=tanh(x) f=y-x*h if (abs(f) -It. 0.000001) go to 1040 fd=-1.0*h-(x/cosh(x)**2) x=x-(f/fd) continue write(*,1050) f ormat ( ' subroutine disp does not converge!!! ') wnum ( i , j ) =x/depth ( i , j ) continue continue continue return end I I I I I I I I 1 I I I t I I I I I APPENDIX n MONTHLY PLOTS OF CLOSURE PROBABILITIES FROM EQUATION (12) FOR EXISTING BATHYMETRY Note that closure probabilities appear as a solid line trace with corresponding scales on the lefthand vertical axis; actual plant inflow rates are dashed-dot-dot traces per the scales on the righthand vertical axis; and dashed traces are required modifications to plant inflow rates to achieve closure probabilities less than 10%. CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE AGUA HEDIONDA LAGOON, CA JULY, 1993 PQ o Ou OH !=> 0.5 0.4-- 0.2 -- 0.1 -- o.o 0 FOR NORMAL OPERATION NORMAL FLOW RATE FLOW RATE for less than 10% CLOSURE PROBABILITY 14 21 TIME ( DAYS ) 28 -- 800.0 --600.0 1000.0 --400.0 0.0 OH O --200.0 < DH CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE PQ<mo K C/Qo-J 1.5 0.4-- 0 AGUA HEDIONDA LAGOON, CA AUGUST, 1993 PROBABILITY FOR NORMAL OPERATION - NORMAL FLOW RATE REDUCED FLOW RATE for less than 10% CLOSURE PROBABILITY 14 21 TIME ( DAYS ) -- 800.0 i-- 600.0 --400.0 1000.0 !-- 200.0 28 W ^ K O &H P-, CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE AGUA HEDIONDA LAGOON, CA SEPTEMBER, 1993 vj.u - K. °'4 - 1— I s o PH tt 0.2- O " 0.1- n n ppnn ATITT TTV T?nr» is.Tn'mtjf ATr KUoArHijl 1 i r UK 1NUKMAL NORMAL FLOW RATE Dpnitr'TPn FT nw PATTT f/^r-KiLiJJUUiiiJj rJLUVY r\AiHi lor ^ ,.-- ;~ v | OPERATION less than 10% CLOSURE PROBABILITY i ': j ; 1 I i j : i i j | j | MMi |i TTfTlI n iii ii n il 1 li;! y I* ^J ~\ iTrft i i Illi il 1 \ 1 ! ti ^—'fl VU\ 1^ :' i ; :MM MM i : i ~N— i i -;• 1;i - 1VJUU.U s — <• - 800.0 TjWJg "^.^- - 600.0 ^ K O- 400.0 ^ |JH i — iE_ - 200.0 < n n 0 14 21 TIME ( DAYS ) 28 CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE AGUA HEDIONDA LAGOON, CA OCTOBER, 1993 CQ <PQ O OH P O CJ 0 XJ.tJ 0.4 - 0.2 - 0.1 - n n _ •pr?nn ARTT T1 KUDArJlLil NORMAL F "DT7*TM Tf Ij1!"^XvHily U v/UjJL/ \\ ; _^_^ TY FOR NORMAL OPERATION 'LOW RATE FLOW RATE for less than 10% CLOSURE PROBABILITY ; / /i ^__ i i i i :^_ i i : i i : i i : : : i ': :' i i :': ;;:'•: : '::!::1 11 Mi ! II !! !! ! 1 L IIi n L^ i —i i 1 ! i ii ^ \ \ \ i Ml\jr~ i y — ii- _ .^. - iwuu. - 800.0 - 600.0 - 400.0 - 200.0 - n n 14 21 TIME ( DAYS ) 28 PH CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE PQ <* PQO PH PH CO OJ O 0.5 0.4 — 0.3 — 0.2 -- 0 AGUA HEDIONDA LAGOON, CA NOVEMBER, 1993 PROBABILITY FOR NORMAL OPERATION - NORMAL FLOW RATE - REDUCED FLOW RATE for less than 10% CLOSURE PROBABILITY -- 800.0 7 14 21 TIME ( DAYS ) 1000.0 -- 200.0 006 _.-- 600.0 O 400.0 J PH CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE PQ PQ O WK C/3 O O 0.4-- 0.2 0.1 -- 0.0 0 AGUA HEDIONDA LAGOON, CA DECEMBER, 1993 PROBABILITY FOR NORMAL OPERATION - NORMAL FLOW RATE FLOW RATE for less than 10% CLOSURE PROBABILITY -- 800.0 14 TIME ( DAYS ) 1000.0 --600.0 --400.0 28 K O &H ^ 200.0 <i—i PH 0.0 CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE AGUA HEDIONDA LAGOON, CA JANUARY, 1994 PQ <PQ O PH w PH CO O 0 7 14 21 TIME ( DAYS ) \J.<J - 0.4 - 0.3 - 0.2- 0.1 - n n _ PROBABILITY FOR NORMAL OPERATION NORMAL FLOW RATE REDUCED FLOW RATE for " -TTTTT ~~' •"1! II!!! ill!•i i 1= i• : i i i i• : „• : i i f ' r ii iiK: i; \l1 I S 1 L^^~^ less than 10% CLOSURE PROBABILITY II !i j || i i j j j i |l ! i i | ! i i j: j i 1! 1! SIS! 1i :ii iiii iiI ii; i J -Mil! i H\• '\ _J ris -r-r'l 111 :1 L : i i i n i _r^" 1UUU. - 800.0 - 600.0 - 400.0 - 200.0 _ n n 28 K O jz;i—i Jz; PH CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE PQ <PQ O D5 PH CO O-J O AGUA HEDIONDA LAGOON, CA FEBRUARY, 1994 0.5 0.4 -- PROBABILITY FOR NORMAL OPERATION - NORMAL FLOW RATE REDUCED FLOW RATE for less than 10% CLOSURE PROBABILITY 0.3 —i 0.2 0.1 -I 0.0 0 -- 800.0 1000.0 600.0 --400.0 -- 200.0 0.0 14 TIME ( DAYS ) 21 28 CO) K ^O CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE PQ<pqo DH CO OnJ O AGUA HEDIONDA LAGOON, CA MARCH, 1994 0.5 0.4-- PROBABILITY FOR NORMAL OPERATION - NORMAL FLOW RATE • REDUCED FLOW RATE for less than 10% CLOSURE PROBABILITY 0.0 0 7 14 21 TIME ( DAYS ) -- 800.0 4-600.0 400.0 1000.0 200.0 5 PH CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE PQ PQ O (X W K OJO 0.5 0.4 -- 0.0 0 AGUA HEDIONDA LAGOON, CA APRIL, 1994 PROBABILITY FOR NORMAL OPERATION - NORMAL FLOW RATE FLOW RATE for less than 10% CLOSURE PROBABILITY 14 TIME ( DAYS ) 21 --800.0 --600.0 400.0 1000.0 200.0 &H OH CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE 0.5 0.4- AGUA HEDIONDA LAGOON, CA MAY, 1994 PROBABILITY FOR NORMAL OPERATION - NORMAL FLOW RATE REDUCED FLOW RATE for less than 10% CLOSURE PROBABILITY 1000.0 -- 800.0 PQ <PQ O PL, C2 uyo 0.2 -- 0.1 -- 0.0 0 PL H= 7 14 TIME ( DAYS ) 28 --600.0 -- 400.0 -- 200.0 0.0 J PH CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE i-Ji— i PP < PQ O OS PL, CO O-J O 0.5 0.4 -- 0.3 -- 0.2 -- AGUA HEDIONDA LAGOON, CA JUNE, 1994 PROBABILITY FOR NORMAL OPERATION - NORMAL FLOW RATE • REDUCED FLOW RATE for less than 10% CLOSURE PROBABILITY 0.1 -- 0 14 21 TIME ( DAYS ) -- 600.0 400.0 1000.0 800.0 4- 200.0 O nJ fciz; p—I 3 PH CLOSURE PROBABILITY DURING BIANNUAL DREDGE CYCLE AGUA HEDIONDA LAGOON, CA JULY, 1994 VJ.U -ROBABILITY0 OCO i^1 1CLOSURE P]3 O O•> »-» roi iC TDTPHP ACTT TTV T?HP NTnTPMAT nPTPTPATTHMr rCUJjAtSlljl 1 I r UK INUKMALi UrlLKAllUlN NORMAL FLOW RATE T? IP TUT PITTI PI nw DAT'IP -F/-.V IQOO tVio-n \ r\°7 PindTTT?!? TDPHRARTT TTVrCrjJJUUrjJJ rLiUW rvAiHj lor less tnan iu/o ULUoUJXiij JrixUDAJDiijii I I' ( — \ /T ..... .... . .1 1 1 1 ) 7 14 21 28 1000.0 -- 800.0 -- 600.0 --400.0 --200.0 0.0 P-H TIME ( DAYS ) I • APPENDIX m OPTIMAL VELOCITY DISTRIBUTIONS FOR STRATFORD • PRESSURE RECOVERY DURING FLOOD FLOW I I I I I I I I I I I I I I I I FLOOD FLOW VELOCITY DISTRIBUTION FOR SPRING TIDE MAXIMUM OCEAN INLET, AGUA HEDIONDA LAGOON, CA CJ <D E-»—i O O OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 90 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) FLOOD FLOW VELOCITY DISTRIBUTION FOR SPRING TIDE MAXIMUM RAILROAD BRIDGE, AGUA HEDIONDA LAGOON, CA o CD eo Oo 200-r 180- 160- 140- 120- 100- 80- 60- 40- 20- 0-- OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 0 30 60 90 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) FLOOD FLOW VELOCITY DISTRIBUTION FOR SPRING TIDE MAXIMUM 1-5 BRIDGE, AGUA HEDIONDA LAGOON, CA o CD00 Oo 200 y 180- 160- 140- 120- 100-- 80- 60- 40- 20- 0- E-2 OPu 2 OCJ o KQ OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 0 20 40 60 80 100 120 140 160 180 200 220 CHANNEL CENTERLINE DISTANCE ( m ) oCDin 6 O Oo FLOOD FLOW VELOCITY DISTRIBUTION FOR NEAP TIDE OCEAN INLET, AGUA HEDIONDA LAGOON, CA OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 30 60 90 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) o CD eo Oo FLOOD FLOW VELOCITY DISTRIBUTION FOR NEAP TIDE RAILROAD BRIDGE, AGUA HEDIONDA LAGOON, CA 50 40- 30- 20- 10- 0 OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 0 30 60 90 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) o CL)w 6 O O O FLOOD FLOW VELOCITY DISTRIBUTION FOR NEAP TIDE 1-5 BRIDGE, AGUA HEDIONDA LAGOON, CA 50 40- 30- 20- 10- 0 Od, eJODi OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 0 20 40 60 80 100 120 140 160 180 200 220 CHANNEL CENTERLINE DISTANCE ( m ) o CD E-i—i O O FLOOD FLOW VELOCITY DISTRIBUTION FOR MEAN TIDE OCEAN INLET, AGUA HEDIONDA LAGOON, CA 120 100-OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 0 30 60 90 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) o <D GO 6 O Oo FLOOD FLOW VELOCITY DISTRIBUTION FOR MEAN TIDE RAILROAD BRIDGE, AGUA HEDIONDA LAGOON, CA 120 110- 100-- 90-- 80- 70-- 60-- 50-- 40-- 30-- 20-- 10-- 0-- OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 0 30 60 90 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) o CD 6 O E-i—i O O FLOOD FLOW VELOCITY DISTRIBUTION FOR MEAN TIDE 1-5 BRIDGE, AGUA HEDIONDA LAGOON, CA 120 100- 80- 60- 40- 20- 0 g Ocu H 2 O CJ O OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 0 20 40 60 80 100 120 140 160 180 200 220 CHANNEL CENTERLINE DISTANCE ( m ) I • APPENDIX IV OPTIMAL VELOCITY DISTRIBUTIONS FOR STRATFORD • PRESSURE RECOVERY DURING EBB FLOWS I I I I I I I I I I I I I I I I o 6 O £-•HHoo t-J EBB FLOW VELOCITY DISTRIBUTION FOR SPRING TIDE MAXIMUM RAILROAD BRIDGE, AGUA HEDIONDA LAGOON, CA 200 OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 0 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) o (D 00 Oo EBB FLOW VELOCITY DISTRIBUTION FOR SPRING TIDE MAXIMUM 1-5 BRIDGE, AGUA HEDIONDA LAGOON, CA 200 H2I—)ODu IH OO OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 20 40 60 80 100 120 140 160 180 200 220 CHANNEL CENTERLINE DISTANCE ( m ) o (D E-h—< Oo EBB FLOW VELOCITY DISTRIBUTION FOR NEAP TIDE RAILROAD BRIDGE, AGUA HEDIONDA LAGOON, CA 50 40 + 30 + 20 + 10 + 0 OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS OPL, 2O 0 30 60 90 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) u CDDO 6 O CJ O EBB FLOW VELOCITY DISTRIBUTION FOR NEAP TIDE 1-5 BRIDGE, AGUA HEDIONDA LAGOON, CA 50 40- 30- 20- 10- 0 O Pu J OKH 55OU O Q>H ffi OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 0 20 40 60 80 100 120 140 160 180 200 220 CHANNEL CENTERLINE DISTANCE ( m ) Oo EBB FLOW VELOCITY DISTRIBUTION FOR MEAN TIDE RAILROAD BRIDGE, AGUA HEDIONDA LAGOON, CA o - CD 6 O 120 -r 110- 100- 90- 80- 70- 60- 50- 40- 30- 20- 10 0 0 OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 30 60 90 120 150 180 210 240 270 300 CHANNEL CENTERLINE DISTANCE ( m ) o CD o EBB FLOW VELOCITY DISTRIBUTION FOR MEAN TIDE 1-5 BRIDGE, AGUA HEDIONDA LAGOON, CA 120 loo-- Bo-- 40- 20- 0 OPTIMAL STRATFORD RECOVERY EXISTING CONDITIONS 100 120 140 160 180 200 220 CHANNEL CENTERLINE DISTANCE ( m ) I • APPENDIX V CALIFORNIA COORDINATES FOR THE FLOW FENCES •I OF THE FLAT BOTTOM SOLUTIONS (STRATFORD FENCE) I I I I I I I I I I I I I I 1 1 California Coordinates 1 1H IH IH 1 1 Ocean X 1665560.000 1665580.000 1665600.000 1665620.000 1665640.000 1665660.000 1665680.000 1665700.000 1665720.000 1665740.000 - 1665760.000 1665780.000 1665800.000 1665820.000 1665840.000 1665860.000 1665880.000 1665900.000 1665920.000 1665940.000 1665960.000 1665980.000 1666000.000 1666020.000 1666040.000 1666060.000 1666080.000 1666100.000 1666120.000 1666140.000 1666160.000 1666180.000 1666200.000 ' 1666220.000 1666240.000 1666260.000 1666280.000 1666300.000 1666320.000 1666340.000 1666360.000 1666380.000 1666400.000 for Flow Fence to Inlet, Right Hand Yn 358036.000 358043.000 358047.000 358053.000 358059.000 358062.000 358065.000 358067.000 358069.000 358070.000 358073.000 358071.000 358071.000 358073.000 358073.000 358073.000 358071.000 358070.000 358070.000 358069.000 358068.000 358065.000 358064.000 358062.000 358061.000 358061.000 358052.000 358043.000 358032.000 358023.000 358008.000 357994.000 357978.000 357960.000 357941.000 357922.000 357894.000 357862.000 357833.000 357803.000 357771.000 357741.000 357712.000 the Flat Bottom Solution: Solution Ys 357860.000 357803.300 357782.800 357769.600 357758.700 357746.400 357734.900 357722.900 357711.500 357699.100 357688.900 357673.700 357660.300 357649.000 357635.300 357621.400 357605.200 357589.500 357574.400 357557.800 357540.500 357520.600 357501.800 357481.300 357460.500 357439.800 357408.600 357375.800 357339.500 357303.000 357258.200 357211.700 357159.500 357101.600 357037.400 . 1 1 1 1 1 1 1^p 1 1 1 1 1 1 1 1 1 I I California Coordinates Railroad X 1667660.000 1667680.000 1667700.000 1667720.000 1667740.000 1667760.000 1667780.000 1667800.000 1667820.000 1667840.000 1667860.000 1667880.000 ' 1667900.000 1667920.000 1667940.000 1667960.000 1667980.000 * for Flow Fence to the Flat Bottom Solution: Bridge, Right Hand Solution Yn 357355.000 357387.200 357404.400 357420.300 357436.100 357452.300 357469.000 357486.900 357506.200 357527.400 357551.200 357578.600 357611.700 357653.000 357709.400 357795.000 357960.900 Ys 357257.000 357259.400 357263.400 357267.600 357271.800 357276.000 357280.100 357284.000 357287.800 357291.400 357294.800 357297.700 357299.900 357301.300 357301.000 357297.400 357284.900 I I I I I I I 1 I I I I I I I I I I I California Coordinates for Flow Fence to the Flat Bottom Solution: Railroad Bridge, Left Hand Solution X 1667660.000 1667640.000 1667620.000 1667600.000 1667580.000 1667560.000 1667540.000 1667520.000 1667500.000 1667480.000 1667460.000 1667440.000 1667420.000 1667400.000 1667380.000 1667360.000 1667340.000 1667320.000 1667300.000 1667280.000 1667260.000 1667240.000 1667220.000 1667200.000 1667180.000 1667160.000 1667140.000 1667120.000 1667100.000 1667080.000 Yn 357355.000 357354.600 357350.000 357345.000 357339.800 357334.400 357329.000 357323.700 357318.300 357313.000 357307.700 357302.500 357297.400 357292.400 357287.400 357282.700 357278.000 357273.600 357269.400 357265.400 357261.700 357258.400 357255.600 357253.300 357251.700 357250.800 357251.300 357253.200 357257.400 357264.900 Ys 357257.000 357233.600 357220.100 357207.600 357195.500 357183.700 357172.000 357160.200 357148.500 357136.600 357124.700 357112.600 357100.300 357087.800 357074.900 357061.800 357048.400 357034.500 357020.100 357005.000 356989.300 356972.800 356955.200 356936.300 356915.800 356893.300 356868.200 356839.300 356805.400 356763.500 I I I I I i i i i i i i i i i i i i D California Coordinates for Flow Fence to the Flat Bottom Solution: 1-5 Bridge, Right Hand Solution 1668600.000 1668620.000 1668640.000 1668660.000 1668680.000 1668700.000 1668720.000 1668740.000 1668760.000 1668780.000 1668800.000 1668820.000 1668840.000 1668860.000 1668880.000 1668900.000 1668920.000 1668940.000 1668960.000 1668980.000 1669000.000 Yn 357465. 357484, 357496. 357506. 357517. 357527. 357537. 357548. 357558. 357569. 357580. 357592. 357605. 357618. 357632. 357648, 357665, 357684, 357707, 357735, 357771, 000 800 200 700 000 300 500 000 600 500 900 700 200 400 600 100 300 800 600 400 500 Ys 357369. 357360. 357359. 357359. 357359. 357360. 357360. 357361. 357361. 357360. 357360. 357359. 357357. 357355. 357351. 357346. 357340. 357331. 357319, 357302. 357276, 000 000 300 400 800 300 800 100 100 900 300 200 400 000 500 700 200 400 300 300 800 1 1 1 1 1 1 I 1 1 1 1 1 1 1 1 1 1 1 1 California Coordinates 1-5 X 1668600.000 1668580.000 1668560.000 1668540.000 1668520.000 1668500.000 1668480.000 1668460.000 1668440.000 1668420.000 1668400.000 1668380.000 * • for Flow Fence to the Flat Bridge, Left Hand Solution Yn 357465.000 357491.000 357499.800 357508.000 357517.100 357528.000 357541.900 357560.200 357585.800 357624.400 357688.600 357824.500 - Bottom Solution: Ys 357369.000 357358.000 357349.000 357340.000 357330.900 357321.600 357311.900 357301.800 357290.900 357278.500 357263.300 357240.100 APPENDIX VI I I CALIFORNIA COORDINATES FOR THE AUXILIARY FLOW FENCES II OF THE VARIABLE DEPTH SOLUTION (STRATFORD BOTTOM) I I I I I I 0 1 1 I I I I I I fl I I I I I I I I I I I I I I I I I I I California Coordinates for Flow Fence to Variable Depth Solution: Ocean Inlet, Right Hand Solution X 1665560.000 1665580.000 1665600.000 1665620.000 1665640.000 1665660.000 1665680.000 1665700.000 1665720.000 1665740.000 1665760.000 1665780.000 1665800.000 1665820.000 1665840.000 1665860.000 - 1665880.000 1665900.000 1665920.000 1665940.000 1665960.000 1665980.000 1666000.000 1666020.000 1666040.000 1666060.000 1666080.000 1666100.000 1666120.000 1666140.000 1666160.000 1666180.000 1666200.000 1666220.000 1666240.000 1666260.000 1666280.000 1666300.000 . 1666320.000 1666340.000 1666360.000 1666380.000 1666400.000 1666420.000 1666440.000 1666460.000 1666480.000 1666500.000 1666520.000 Yn 357948.000 357954.000 357958.500 357961.500 357963.500 357963.500 357961.500 357958.500 357955.000 357938.000 357925.500 357909.500 357906.000 357875.500 357859.000 357842.000 357825.000 357809.000 357794.000 357779.000 357764.500 357748.000 357734.000 357719.500 357706.500 357691.500 357674.000 357658.000 357640.500 357624.500 357605.500 357587.000 357566.000 357543.500 357519.000 357495.000 357466.500 357434.500 357403.500 357370.500 357337.000 357305.000 357273.500 357239.000 357206.000 357177.000 357154.000 357135.500 357106.500 YS 357860.00 357865.00 357870.00 357870.00 357868.00 357865.00 357858.00 357850.00 357841.00 357806.00 357778.00 357748.00 357741.00 357678.00 357645.00 357611.00 357579.00 357548.00 357518.00 357489.00 357461.00 357431.00 357404.00 357377.00 357352.00 357322.00 357296.00 357273.00 357249.00 357226.00 357203.00 357180.00 I I I I I I I I I I I I I I I I I I I California Coordinates for Flow Fence to Variable Depth Solution: Railroad Bridge, Right Hand Solution X 1667660. 1667680. 1667700. 1667720. 1667740. 1667760. 1667780. 1667800. 1667820. 1667840. 1667860. 1667880. 1667900. 1667920. 1667940. 1667960. 1667980. 1668000, 000 000 000 000 000 000 000 000 000 000 000 000 000 000 000 000 000 000 Yn 357306. 357320. 357770. 357775. 357779. 357781. 357782. 357782. 357783. 357784. 357786. 357790. 357795. 357798. 357801. 357805. 357801, 357806. 000 000 500 500 500 500 000 000 500 000 500 000 500 500 500 000 000 500 YS 357257.000 357269.000 357252.100 357255.400 357258.700 357261.900 357264.900 357267.800 357270.400 357272.600 357274.300 357275.200 357275.000 357273.200 357268.300 357257.600 357282.000 357294.000 I I I I I I I I I I California Coordinates for Flow Fence to Variable Depth Solution: Railroad Bridge, Left Hand Solution I I I I I I I I X 1667660 1667640 1667620 1667600 1667580 1667560 1667540 1667520 1667500 1667480 1667460 1667440 1667420 1667400 1667380 1667360 1667340 1667320 1667300 1667280 1667260 1667240 1667220 1667200 1667180 1667160 1667140 1667120 1667100 1667080 1667060 1667040 1667020 1667000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 .000 Yn 357306.000 357295.000 357284.500 357268.500 357261.500 357250.500 357230.000 357229.000 357221.000 357209.000 357198.000 357177.000 356779.000 356749.000 356720.500 356679.000 356373.000 356376.000 356376.000 356370.500 356360.500 356350.500 356357.000 356365.000 356373.500 356383.500 356394.500 356403.500 356425.000 356495.000 356507.000 356520.000 356534.500 356551.000 Ys 357257.000 357245.000 357232.000 357217.000 357204.000 357192.000 357179.000 357167.000 357158.000 357142.000 357128.000 357093.000 356887.300 356863.600 356839.300 356814.300 356788.400 356761.500 356733.500 356704.000 356673.000 356640.000 356604.500 356566.200 356524.000 I I I I I I 1 I I I I I I I I I I I I California Coordinates for Flow Fence to Variable Depth Solution: 1-5 Bridge, Right Hand Solution 1668600.000 1668620.000 1668640.000 1668660.000 1668680.000 1668700.000 1668720.000 1668740.000 1668760.000 1668780.000 1668800.000 1668820.000 1668840.000 1668860.000 1668880.000 1668900.000 1668920.000 1668940.000 1668960.000 1668980.000 1669000.000 1669020.000 1669040.000 1669060.000 Yn 357417. 357418. 357417. 357418. 357419. 357422. 357429. 357558. 357893. 357880. 357867. 357857. 357847. 357831. 357801. 357769. 357748. 357731. 357722, 357712. 357698, 357685, 357673, 357667, 000 000 500 500 500 000 000 000 500 000 500 000 000 500 000 000 000 000 000 000 500 000 500 500 Ys 357369. 357370. 357368. 357369. 357371. 357371. 357359. 357239. 357229. 357218. 357207. 357195. 357181. 357166. 357149. 357130. 357108. 357081. 357048. 357005. 356947, 000 000 000 000 000 000 000 400 400 800 400 200 600 800 900 600 200 200 100 600 000 I I I I I I I I I I I I I I I I I California Coordinates for Flow Fence to Variable Depth Solution: 1-5 Bridge, Left Hand Solution X Yn Ys 1668600.000 357417.000 357369.000 1668580.000 357415.500 357367.000 1668560.000 357414.500 357364.000 1668540.000 357419.000 357361.000 1668520.000 357427.500 357358.000 1668500.000, 357449.500 357355.000 1668480.000 357476.000 357347.000 1668460.000 357490.500 357329.000 1668440.000 357528.500 357316.000 1668420.000 357698.500 357312.900 1668400.000 357744.000 357297.800 1668380.000 357778.500 357310.000 1668360.000 357814.000 357309.000 I || APPENDIX VH I WATERLINES AND EAST BASIN WATER ELEVATIONS FOR STAGE 1 DREDGE PLAN DURING EXTREME SPRING AND NEAP TIDES AND MEAN TIDE CONDITIONS (Plant Flow Rate = 570 mgd) I I I I I I 1 I I I I I 1 1 Zm = +3.05 ft NGVD DREDGE PLAN: STAGE 1 HYDRODYNAMIC TIDE SIMULATION EXTREME SPRING TIDE HHW TOTAL WETTED AREA = 12.191,386 ftz TIDAL PRISM = 69,518,492 ft3 INTER TIDAL AREA = 101.6 ACRES 500 1000 1500 3000 3500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.24 ft HEAD LOSS = 0.39 ft P DR SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL = -4.45 ft NGVD = -4.04 ft NGVD HEAD LOSS = 0.13 ft HEAD LOSS = 0.41 ft DREDGE PLAN: STAGE 1 HYDRODYNAMIC TIDE SIMULATION EXTREME SPRING TIDE LLW TOTAL TYETTED AREA = 7,766.951 ft; TIDAL PRISM = 69,518,492 ft3 SUB TIDAL AREA = 178.3 ACRES 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET j = -4.04 ft NGVD Z, = -4.58 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL MAXIMUM SPRING TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN STAGE 1 DREDGE PLAN: AGUA HEDIONDA LAGOON, CA 5 Q>O iz;o t 4 3 2 1 0 -1 -2 -3 -4 OCEAN TIDE EAST BASIN 8 14 20 26 32 TIME ( hrs ) 38 44 Zm = +2.09 ft NGVD HEAD LOSS = 0.83 ft DREDGE PLAN: STAGE 1 HYDRODYNAMIC TIDE SIMULATION MEAN TIDE HHW TOTAL WETTED AREA = 11,646.075 fl' TIDAL PRISM = 42,452,387 ft3 INTERTIDAL AREA = 35.5 ACRES 0 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.24 ft HEAD LOSS = 0.33 ft Zm = +1.76 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD &: JOSEPH WASYL = -2.23 ft NGVD i = -1.92 ft NGVD DREDGE PLAN: STAGE 1 HYDRODYNAMIC TIDE SIMULATION MEAN TIDE LLW TOTAL "WETTED AREA = 10,100.923 ft'* TIDAL PRISM = 42,452,387 ft3 SUB TIDAL AREA = 232 ACRES 0 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.12 ft HEAD LOSS = 0.31 ft i = -1.92 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL Q>Oiz; MEAN TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN STAGE 1 DREDGE PLAN: AGUA HEDIONDA LAGOON, CA -3-- _4~ OCEAN TIDE EAST BASIN 8 14 20 26 TIME ( hrs ) 32 38 44 Zm = +0.58 ft NGVD HEAD LOSS = 0.51 ft i DREDGE PLAN: STAGE 1 HYDRODYNAMIC TIDE SIMULATION EXTREME NEAP TIDE HHW TOTAL WETTED AREA = 11,219.251 ftj TIDAL PRISM = 22,640,407 ft3 INTERTIDAL AREA = 22.7 ACRES 500 1000 1500 3000 3500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.18 ft 7m = +0.71 ft' NGVD Zm = +0.40 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL = -1.79 ft NGVD DREDGE PLAN: STAGE 1 HYDRODYNAMIC TIDE SIMULATION EXTREME NEAP TIDE LLW TOTAL WETTED AREA = 10.230,865 ft' TIDAL PRISM - 22.640,407 ft3 SUB TIDAL AREA = 234 ACRES 500 1000 1500 eooo HORIZONTAL SCALE IN FEET Z500 HEAD LOSS = 0.10 ft i = -1.57 ft NGVD HEAD LOSS = 0.22 ft = -1.57 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL Q> O o PS NEAP TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN STAGE 1 DREDGE PLAN: AGUA HEDIONDA LAGOON, CA OCEAN TIDE EAST BASIN 8 20 26 TIME ( hrs ) 38 44 I I I APPENDIX I I i WATERLINES AND EAST BASIN WATER ELEVATIONS FOR STAGE 2 DREDGE PLAN DURING EXTREME SPRING AND NEAP TIDES AND MEAN TIDE CONDITIONS (Plant Flow Rate = 570 mgd) I I I I I I I I I I I Zm = +3.05 ft NGVD DREDGE PLAN: STAGE 2 HYDRODYNAMIC TIDE SIMULATION EXTREME SPRING TIDE HHW TOTAL WETTED AREA = 12.311,294 ftj TIDAL PRISM = 72,412,713 ft3 INTERTIDAL AREA = 106.2 ACRES 500 1000 1500 2000 HORIZONTAL SCALE IN FEET Z500 HEAD LOSS = 0.24 ft HEAD LOSS = 0.97 ft HEAD LOSS = 0.22 ft Zm = +3.29 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL = -4.45 ft NGVD = -4.24 ft NGVD HEAD LOSS - 0.13 ft HEAD LOSS = 0.21 ft DREDGE PLAN: STAGE 2 HYDRODYNAMIC TIDE SIMULATION EXTREME SPRING TIDE LLW TOTAL WETTED AREA = 7,684,119 ft TIDAL PRISM = 72,412,713 ft3 SUB TIDAL AREA = 176.4 ACRES 500 1000 1500 3000 HORIZONTAL SCALE IN FEET 2500 j = -4.24 ft NGVD Z. ='-4.58 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL MAXIMUM SPRING TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN STAGE 2 DREDGE PLAN: AGUA HEDIONDA LAGOON, CA 5 Q>O 4 3 2 1 0 -1 -2 -3 _4~ -5 8 OCEAN TIDE EAST BASIN 14 20 26 32 TIME ( hrs ) 38 44 i Zm = +2.09 ft NGVD HEAD LOSS = 0.83 ft , DREDGE PLAN: STAGE 2 HYDRODYNAMIC TIDE SIMULATION MEAN TIDE HHW TOTAL WETTED AREA = 11,675,754 ft! TIDAL PRISM = 45,085,563 ft3 INTERTIDAL AREA = 38.3 ACRES 0 500 1000 1500 3000 3500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.24 ft HEAD LOSS = 0.18 ft Zm = +1.93 ft NGVD DR. SCOTT A JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD &: JOSEPH WASYL = -2.23 ft NGVD -2.09 ft NGVD DREDGE PLAN: STAGE 2 HYDRODYNAMIC TIDE SIMULATION MEAN TIDE LLW TOTAL WETTED AREA = 10.008,132 ftj TIDAL PRISM - 45,085,563 ft3 SUB TIDAL AREA = 229 ACRES 0 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.12 ft HEAD LOSS - 0.14 ft = -2.09 ft NGVD DR. SCOTT A JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL Q>Oiz; iz;o MEAN TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN STAGE 2 DREDGE PLAN: AGUA HEDIONDA, CA 14 OCEAN TIDE EAST BASIN 20 26 TIME ( hrs ) 32 38 44 Zm = +0.58 ft NGVD HEAD LOSS = 0.51 ft DREDGE PLAN: STAGE 2 HYDRODYNAMIC TIDE SIMULATION EXTREME NEAP TIDE HHW TOTAL TOTTED AREA = 11,244,285 ft£ TIDAL PRISM = 24,331,471 ft3 INTERTIDAL AREA = 23.7 ACRES 500 1000 1500 3000 HORIZONTAL SCALE IN FEET 2500 HEAD LOSS = 0.13 ft HEAD LOSS « 0.09 ft Zm = +0.71 ft NGVD Zm = +0.49 ft NGVD DR. SCOTT A JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL = -1.79 ft NGVD DREDGE PLAN: STAGE 2 HYDRODYNAMIC TIDE SIMULATION EXTREME NEAP TIDE LLW TOTAL WETTED AREA = 10,211,747 ft: TIDAL PRISM = 24.331,471 ft3 SUB TIDAL AREA = 234 ACRES 500 1000 1500 2000 HORIZONTAL SCALE IN FEET 2500 HEAD LOSS = 0.10 ft i = -1.69 ft NGVD HEAD LOSS = 0.10 ft Z* = -1.89 ft NGVD = -1.69 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL a •* • Oi—iE- NEAP TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN STAGE 2 DREDGE PLAN: AGUA HEDIONDA LAGOON, CA OCEAN TIDE EAST BASIN 14 20 26 TIME ( hrs ) 32 38 44 § APPENDIX IX I WATERLINES AND EAST BASIN WATER ELEVATIONS FOR STAGE 3 DREDGE PLAN DURING EXTREME SPRING AND NEAP TIDES AND MEAN TIDE CONDITIONS (Plant Flow Rate = 570 mgd) I I I I I I I I I I i t I I I Zm = +3.01 ft NGVD HEAD LOSS = ' +0.99 ft DREDGE PLAN: STAGE 3 HYDRODYNAMIC TIDE SIMULATION EXTREME SPRING TIDE HHW TOTAL WETTED AREA = 12,053,785 ft£ TIDAL PRISM = 76,834,563 ft3 INTERTIDAL AREA = 61.8 ACRES 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.26 ft HEAD LOSS = 0.24 ft Zm - +3.27 ft NGVD Zm = +2.77 ft NGVD DR. SCOTT A JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL Zm = -4.44 ft NGVD DREDGE PLAN: STAGE 3 HYDRODYNAMIC TIDE SIMULATION EXTREME SPRING TIDE LLW TOTAL WETTED AREA = 9,361,328 ft2 TIDAL PRISM = 76.834,563 ft3 SUBTIDAL AREA = 214.9 ACRES 500 1000 1500 2000 HORIZONTAL SCALE IN FEET 2500 HEAD LOSS = 0.14 ft Zm = -4.22 ft NGVD HEAD LOSS = 0.22 ft Zm 4.22 ft NGVD DR. SCOTT A JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL I MAXIMUM SPRING TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN STAGE 3 DREDGE PLAN: AGUA HEDIONDA LAGOON, CA 5 o52; •»-> iz;o I—H I OCEAN TIDE EAST BASIN -5 14 20 26 TIME ( hrs ) 32 38 44 Zm = +2.09 ft NGVD HEAD LOSS 0.83 ft Zm = +1.93 ft NGVD DREDGE PLAN: STAGE 3 HYDRODYNAMIC TIDE SIMULATION MEAN TIDE HHW TOTAL JtETTED AREA m 11,665.127 fts TIDAL PRISM = 45,715,523 ft3 INTERTOAL AREA « 30.8 ACRES 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET Z™ = +1.93 ft NGVD HEAD LOSS = 0.18 ft DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL Zm = -2.23 ft NGVD DREDGE PLAN: STAGE 3 HYDRQDYNAMIC TIDE SIMULATION MEAN TIDE LLW TOTAL WETTED AREA * 10.321.740 ft£ TIDAL PRISM = 45.715,523 ft3 SUB TIDAL AREA = 236.9 ACRES 900 1000 1500 2000 HORIZONTAL SCALE IN FEET 2500 HEAD LOSS = 0.12 ft Zm = -2.09 ft NGVD HEAD LOSS = 0.14 ft Zm = -2.35 ft NGVD Zm = -2.09 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL Q>Oiz; MEAN TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN STAGE 3 DREDGE PLAN: AGUA HEDIONDA LAGOON, CA -3-- -4-- -5 - OCEAN TIDE EAST BASIN 8 14 20 26 TIME ( hrs ) 32 38 44 = +0.58 ft NGVD HEAD LOSS = 0.51 ft DREDGE PLAN: STAGE 3 HYDRODYNAMIC TIDE SIMULATION EXTREME NEAP TIDE HHW TOTAL WETTED AREA = 11,278.666 ft2 TIDAL PRISM = 24,625,587 ft3 INTERTIDAL AREA = 19.3 ACRES 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.13 ft Zm = +0.49 ft NGVD HEAD LOSS = 0.09 ft Zm = +0.71 ft NGVD Zm = +0.49 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL Zm = -1.79 ft NGVD DREDGE PLAN: STAGE 3 HYDRODYNAMIC TIDE SIMULATION EXTREME NEAP TIDE LLW TOTAL WETTED AREA = 10,436,725 TIDAL PRISM = 24,625,587 ft3 SUBTIDAL AREA = 239.6 ACRES 0 500 1000 1500 8000 2500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.10 ft Zrr, - -1.69 ft NGVD HEAD LOSS = 0.10 ft Zm = -1.89 ft NGVD Zm = -1.69 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL Q>O NEAP TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN STAGE 3 DREDGE PLAN: AGUA HEDIONDA LAGOON, CA OCEAN TIDE EAST BASIN 20 26 TIME ( hrs ) 38 44 I I APPENDIX X WATERLINES AND EAST BASIN WATER ELEVATIONS FOR STAGE 4 DREDGE PLAN * DURING EXTREME SPRING AND NEAP TIDES AND MEAN TIDE CONDITIONS (Plant Flow Rate = 570 mgd) I I I I I t I I I I I I I I I I Zm = +3.67 ft NGVD DREDGE PLAN: STAGE 4 HYDRODYNAMIC TIDE SIMULATION EXTREME SPRING TIDE HHW TOTAL WETTED AREA = 12,826,411 ftj TIDAL PRISM = 88.249,822 ft3 INTERTIDAL AREA = 71.5 ACRES 0 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.23 ft Zm = +3.43 ft NGVD HEAD LOSS = 0.24 ft Zm = +3.43 ft NGVD DR. SCOTT A JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL Zt = -4.44 ft NGVD Zt = -4.58 ft NGVD DREDGE PLAN: STAGE 4 HYDRODYNAMIC TIDE SIMULATION EXTREME SPRING TIDE LLW TOTAL WETTED AREA = 9,713,639 ft* TIDAL PRISM = 88,249,822 ft3 SUBTIDAL AREA = 222.9 ACRES 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.14 ft i = -4.22 ft NGVD HEAD LOSS = 0.22 ft i = -4.22 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL MAXIMUM SPRING TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN STAGE 4 DREDGE PLAN: AGUA HEDIONDA LAGOON, CA 5 Q>Oiz; OCEAN TIDE EAST BASIN 14 20 26 TIME ( hrs ) 32 38 44 Zm = +2.67 ft NGVD DREDGE PLAN: STAGE 4 HYDRODYNAMIC TIDE SIMULATION MEAN TIDE HHW TOTAL WETTED AREA = 12,016,143 ft2 TIDAL PRISM = 52,678,459 ft3 INTERTIDAL AREA = 38.9 ACRES 500 1000 1500 2000 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.20 ft HEAD LOSS = 0.29 ft Zm = +2.49 ft NGVD HEAD LOSS = 0.18 ft Zm = +2.87 ft NGVD 3500 Zm = +2.49 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL r = -2.23 ft NGVD DREDGE PLAN: STAGE 4 HYDRODYNAMIC TIDE SIMULATION MEAN TIDE LLW TOTAL WETTED AREA =. 10,321,829 ft£ TIDAL PRISM = 52,678,459 ft3 SUBTIDAL AREA = 236.9 ACRES 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.12 ft HEAD LOSS = 0.14 ft = -2.09 ft NGVD tilt SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL o MEAN TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN STAGE 4 DREDGE PLAN: AGUA HEDIONDA LAGOON, CA 3 -- -4-- -5 8 OCEAN TIDE EAST BASIN 14 20 26 TIME ( hrs ) 32 38 44 Zm = +0.87 ft NGVD Zm = +0.99 ft NGVD DREDGE PLAN: STAGE 4 HYDRODYNAMIC TIDE SIMULATION EXTREME NEAP TIDE HHW TOTAL WETTED AREA =• 11,305.766 ft* TIDAL PRISM = 27.594,738 ft3 INTERTIDAL AREA = 20.4 ACRES 0 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.09 ft Zm = +0.78 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL -1.79 ft NGVD DREDGE PLAN: STAGE 4 HYDRODYNAMIC TIDE SIMULATION EXTREME NEAP TIDE LLW TOTAL WETTED AREA = 10,417.475 ft2 TIDAL PRISM = 27,594.738 ft3 SUBTIDAL AREA = 239.1 ACRES 500 1000 1500 2000 2500 HORIZONTAL SCALE IN FEET HEAD LOSS = 0.10 ft HEAD LOSS = 0.10 ft -1.69 ft NGVD DR. SCOTT A. JENKINS CONSULTING AGUA HEDIONDA SCOTT A. JENKINS PhD & JOSEPH WASYL Q> O Oi— i E- W NEAP TIDE WATER ELEVATIONS: OCEAN vs EAST BASIN STAGE 4 DREDGE PLAN: AGUA HEDIONDA LAGOON, CA OCEAN TIDE EAST BASIN 14 20 26 TIME ( hrs ) 32 38 44